pdf_AISC Design Guide 07-3rd ED 2019.pdf

Design Guide 7
Industrial
Building
Design
Third Edition
https://t.me/Seismic_Control

American Institute of Steel Construction
James M. Fisher, PE, PhD
Design Guide 7
Industrial
Building
Design
Third Edition

© AISC 2019
by
All rights reserved. This book or any part thereof must not be reproduced
in any form without the written permission of the publisher.
The AISC logo is a registered trademark of AISC.
The information presented in this publication has been prepared following recognized principles of design
and construction. While it is believed to be accurate, this information should not be used or relied upon
for any specific application without competent professional examination and verification of its accuracy,
suitability and applicability by a licensed engineer or architect. The publication of this information is not a
representation or warranty on the part of the American Institute of Steel Construction, its officers, agents,
employees or committee members, or of any other person named herein, that this information is suitable
for any general or particular use, or of freedom from infringement of any patent or patents. All represen-
tations or warranties, express or implied, other than as stated above, are specifically disclaimed. Anyone
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Caution must be exercised when relying upon standards and guidelines developed by other bodies and
incorporated by reference herein since such material may be modified or amended from time to time sub-
sequent to the printing of this edition. The American Institute of Steel Construction bears no responsibility
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of this edition.
Printed in the United States of America

iii
Author
James M. Fisher, Ph.D., P.E., DIST.M.ASCE, is Vice President Emeritus of CSD Engineers. He is a member of the AISC Com-
mittee on Specifications and its task committee on Member Design and Stability. He also serves as chairman of the committee
developing the Seismic Provisions for Evaluation and Retrofit of Structural Steel Buildings, AISC 342.
Acknowledgments
The author thanks the American Iron and Steel Institute for funding the first edition of this Guide and the American Institute of
Steel Construction for funding the second and third editions. The author also appreciates the guidance from the AISC review
committee and staff members who contributed many suggestions:
Steve Bohm
Eric Bolin
Cynthia Duncan
Lou Geschwindner
Roger LaBoube
Steve Herlache
Pete Cheever
Larry Kloiber
Larry Kruth
Margaret Matthew
Curt Miller
Preface
This Design Guide provides guidance for the design of both light and heavy industrial buildings. As in previous editions, build-
ings with and without overhead cranes are discussed. The third edition of this Design Guide incorporates the 2016 AISC Specifi-
cation and the 15th Edition of the AISC Steel Construction Manual. Analysis and design examples are provided in greater detail
than in the previous editions of the Design Guide.

iv

v
Table of Contents
INTRODUCTION . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 1
PART 1 INDUSTRIAL BUILDINGS—
GENERAL . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 3
CHAPTER 1 LOADING CONDITIONS AND
LOAD COMBINATIONS . . . . . . . . . . . . . . . . . . 3
CHAPTER 2 OWNER-ESTABLISHED
CRITERIA . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 5
2.1 SLABS-ON-GROUND DESIGN . . . . . . . . . . .
5
2.2 JIB CRANES . . . . . . . . . . . . . . . . . . . . . . . .
5
2.3 INTERIOR VEHICULAR TRAFFIC . . . . . . . . .
5
2.4 FUTURE EXPANSION . . . . . . . . . . . . . . . . .
6
2.5 DUST CONTROL/EASE
OF MAINTENANCE . . . . . . . . . . . . . . . . . . .
6
2.6 ELECTRICAL, PIPING, AND
EQUIPMENT LOADS . . . . . . . . . . . . . . . . . .
6
CHAPTER 3 ROOF SYSTEMS . . . . . . . . . . . . . . . . . . 7
3.1 STEEL DECK FOR BUILT-UP OR
MEMBRANE ROOFS . . . . . . . . . . . . . . . . . .
7
3.2 METAL ROOFS . . . . . . . . . . . . . . . . . . . . . .
7
3.3 INSULATION AND ROOFING . . . . . . . . . . . .
9
3.4 EXPANSION JOINTS . . . . . . . . . . . . . . . . . .
9
3.5 ROOF PITCH, DRAINAGE, AND PONDING . 11
3.6 JOISTS AND PURLINS . . . . . . . . . . . . . . . . 12
3.7 ROOF PENETRATIONS AND EQUIPMENT . . 13
CHAPTER 4 ROOF TRUSSES . . . . . . . . . . . . . . . . . 15
4.1 GENERAL DESIGN AND
ECONOMIC CONSIDERATIONS . . . . . . . . . 15
4.2 CONNECTION CONSIDERATIONS . . . . . . . 16
4.3 TRUSS BRACING . . . . . . . . . . . . . . . . . . . .
16
4.3.1 Roof Truss Stability Bracing Example . . 17
4.4 ERECTION BRACING . . . . . . . . . . . . . . . . 19
4.5 OTHER CONSIDERATIONS . . . . . . . . . . . . 20
CHAPTER 5 WALL SYSTEMS . . . . . . . . . . . . . . . . . 21
5.1 FIELD-ASSEMBLED PANELS . . . . . . . . . . .
21
5.2 FACTORY-ASSEMBLED PANELS . . . . . . . . 21
5.3 PRECAST WALL PANELS . . . . . . . . . . . . . .
22
5.4 MASONRY WALLS . . . . . . . . . . . . . . . . . . 23
5.5 GIRTS . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 23
5.6 WIND COLUMNS . . . . . . . . . . . . . . . . . . . 24
CHAPTER 6 FRAMING SCHEMES . . . . . . . . . . . . 27
6.1 BRACED FRAMES VERSUS
RIGID FRAMES . . . . . . . . . . . . . . . . . . . . .
27
6.2 HSS VERSUS W-SHAPE COLUMNS . . . . . . .27
6.3 MEZZANINE AND PLATFORM FRAMING . . 27
6.4 ECONOMIC CONSIDERATIONS . . . . . . . . .
28
CHAPTER 7 BRACING SYSTEMS . . . . . . . . . . . . . 31
7.1 RIGID-FRAME SYSTEMS . . . . . . . . . . . . . . 31
7.2 BRACED SYSTEMS . . . . . . . . . . . . . . . . . . 31
7.2.1 Roof Diaphragms . . . . . . . . . . . . . . . .
31
7.2.1.1 Diaphragm Design Example . . . . . . . . . 32
7.2.2 Roof X-Bracing . . . . . . . . . . . . . . . . . 36
7.2.3 Vertical Bracing . . . . . . . . . . . . . . . . .
36
7.3 TEMPORARY BRACING . . . . . . . . . . . . . . 36
CHAPTER 8 COLUMN ACHORAGE . . . . . . . . . . . 39
8.1 RESISTING TENSILE FORCES WITH
ANCHOR RODS . . . . . . . . . . . . . . . . . . . . .
39
8.2 PARTIAL BASE FIXITY . . . . . . . . . . . . . . . 40
CHAPTER 9 SERVICEABILITY CRITERIA . . . . . 41
9.1 SERVICEABILITY CRITERIA FOR
ROOF DESIGN . . . . . . . . . . . . . . . . . . . . . . 41
9.2 METAL WALL PANELS . . . . . . . . . . . . . . . 41
9.3 PRECAST WALL PANELS . . . . . . . . . . . . . . 42
9.4 MASONRY WALLS . . . . . . . . . . . . . . . . . . 42
PART 2 INDUSTRIAL BUILDINGS
WITH CRANES . . . . . . . . . . . . . . . . . . . . . . . . . 43
CHAPTER 10 INTRODUCTION TO PART 2 . . . . . 43
10.1 AIST TR-13 BUILDING
CLASSIFICATIONS . . . . . . . . . . . . . . . . . . 43
10.2 CMAA CRANE CLASSIFICATIONS . . . . . . .
43
CHAPTER 11 FATIGUE . . . . . . . . . . . . . . . . . . . . . . . 47
11.1 FATIGUE DAMAGE . . . . . . . . . . . . . . . . . . 47
11.2 CRANE RUNAWAY FATIGUE
CONSIDERATIONS . . . . . . . . . . . . . . . . . . 48
CHAPTER 12 CRANE-INDUCED LOADS AND
LOAD COMBINATIONS . . . . . . . . . . . . . . . . . 53
12.1 VERTICAL IMPACT . . . . . . . . . . . . . . . . . . 53
12.2 SIDE THRUST . . . . . . . . . . . . . . . . . . . . . .
53

vi
12.3 LONGITUDINAL OR TRACTIVE FORCE . . . 54
12.4 CRANE STOP FORCES . . . . . . . . . . . . . . . .
54
12.5 ECCENTRICITIES . . . . . . . . . . . . . . . . . . . 55
12.6 SEISMIC LOADS . . . . . . . . . . . . . . . . . . . . 55
12.7 LOAD COMBINATIONS . . . . . . . . . . . . . . . 55
CHAPTER 13 STRUCTURAL SYSTEMS IN
CRANE BUILDINGS . . . . . . . . . . . . . . . . . . . . . 57
13.1 ROOF SYSTEMS . . . . . . . . . . . . . . . . . . . . 57
13.2 WALL SYSTEMS . . . . . . . . . . . . . . . . . . . . 57
13.3 FRAMING SYSTEMS . . . . . . . . . . . . . . . . . 57
13.4 BRACING SYSTEMS . . . . . . . . . . . . . . . . . 58
13.4.1 Roof Bracing . . . . . . . . . . . . . . . . . . 58
13.4.2 Wall Bracing . . . . . . . . . . . . . . . . . . . 59
CHAPTER 14 CRANE RUNWAY DESIGN . . . . . . . 61
14.1 CRANE RUNWAY BEAM
DESIGN PROCEDURE . . . . . . . . . . . . . . . . 61
14.1.1 Crane Runway Girder
Design Example (ASD) . . . . . . . . . . . .
63
14.1.2 Crane Runway Girder
Design Example (LRFD) . . . . . . . . . . .
68
14.1.3 Crane Runway Girder with Cap Channel
Design Example (ASD) . . . . . . . . . . . .
70
14.1.4 Crane Runway Girder with Cap Channel
Design Example (LRFD) . . . . . . . . . . .
75
14.2 PLATE GIRDERS . . . . . . . . . . . . . . . . . . . . 77
14.3 SIMPLE SPAN VERSUS
CONTINUOUS RUNWAYS . . . . . . . . . . . . . 77
14.4 LATERAL LOAD-RESISTING MEANS . . . . . 80
14.4.1 Cap Channels, Cap Plates, or Angles
Welded to the Top Flange . . . . . . . . . . 80
14.4.2 Oversized Top Flange . . . . . . . . . . . . . 80
14.4.3 Backup Trusses and Apron Plates . . . . . 80
14.5 RUNWAY BRACING CONCEPTS . . . . . . . . . 80
14.6 CRANE STOPS . . . . . . . . . . . . . . . . . . . . . 82
14.7 CRANE RAIL ATTACHMENTS . . . . . . . . . . 82
14.7.1 Hook Bolts . . . . . . . . . . . . . . . . . . . . 82
14.7.2 Rail Clips . . . . . . . . . . . . . . . . . . . . . 82
14.7.3 Rail Clamps . . . . . . . . . . . . . . . . . . . 82
14.7.4 Patented Rail Clips . . . . . . . . . . . . . . . 83
14.7.5 Design of Rail Attachments . . . . . . . . . 83
14.7.6 Rail Pads . . . . . . . . . . . . . . . . . . . . . 83
14.8 CRANE RAILS AND CRANE
RAIL JOINTS . . . . . . . . . . . . . . . . . . . . . . .
84
14.9 RUNWAY CLEARANCES, TOP OF
RAIL ELEVATION, AND
BUILDING EAVE HEIGHT . . . . . . . . . . . . . 84
CHAPTER 15 CRANE RUNWAY FABRICATION
AND ERECTION TOLERANCES . . . . . . . . . . 85
CHAPTER 16 COLUMN DESIGN . . . . . . . . . . . . . . 87
16.1 BASE FIXITY AND LOAD SHARING . . . . . . 87
16.2 PRELIMINARY DESIGN METHODS . . . . . . 92
16.2.1 Stepped Columns . . . . . . . . . . . . . . . . 92
16.2.2 Double Columns (Laced or Tied) . . . . . 94
16.2.3 Single Columns (Bracketed) . . . . . . . . .
94
16.3 FINAL DESIGN PROCEDURES . . . . . . . . . . 92
16.3.1 Bracketed Crane Column
Design Example . . . . . . . . . . . . . . . . 97
16.3.2 Stepped Crane Column
Design Example . . . . . . . . . . . . . . . 101
16.4 ECONOMIC CONSIDERATIONS . . . . . . . . 106
CHAPTER 17 OTHER CRANE
CONSIDERATIONS . . . . . . . . . . . . . . . . . . . . 109
17.1 OUTSIDE CRANES . . . . . . . . . . . . . . . . . 109
17.2 UNDERHUNG CRANES . . . . . . . . . . . . . . 109
17.3 MAINTENANCE AND REPAIR . . . . . . . . . 112
CHAPTER 18 SUMMARY AND
DESIGN PROCEDURES . . . . . . . . . . . . . . . . . 113
APPENDIX . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 115
REFERENCES . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 117

AISC DESIGN GUIDE 7 / INDUSTRIAL BUILDING DESIGN / 1
Although the basic structural and architectural components
of industrial buildings are relatively simple, combining all
of the elements into a functional economical building can
be a complex task. Criteria to accomplish this task can be
stated. The purpose of this Guide is to provide the indus-
trial building designer with guidelines and design criteria for
the design of buildings without cranes or for buildings with
light-to-medium duty cycle cranes. Part 1 deals with gen-
eral topics on industrial buildings. Part 2 deals with struc-
tures containing cranes. Requirements for seismic detailing
for industrial buildings have not been addressed in this
Guide. Any special detailing for seismic conditions must be
addressed by the designer.
Most industrial buildings primarily serve as an enclo-
sure for production and/or storage. The design of industrial
buildings may seem logically the province of the structural
engineer. It is essential to realize that most industrial build-
ings involve much more than structural design. The designer
may assume an expanded role and may be responsible for
site planning, establishing grades, handling surface drain-
age, parking, onsite traffic, building aesthetics, and perhaps,
landscaping. Access to rail and the establishment of proper
floor elevations (depending on if direct fork truck entry to
rail cars is required) are important considerations. Proper
clearances to sidings and special attention to curved siding
and truck grade limitations are also essential.
Introduction

2 / INDUSTRIAL BUILDING DESIGN / AISC DESIGN GUIDE 7

PART 1
INDUSTRIAL BUILDINGS—GENERAL
Chapter 1
Loading Conditions and Load Combinations
Loading conditions and load combinations for industrial
buildings without cranes are well established by building
codes.
Loading conditions are categorized as follows:
1. Dead load: This load represents the weight of the
structure and its components and is usually expressed
in pounds per square foot. In an industrial building,
the building use and industrial process usually involve
permanent equipment that is supported by the struc-
ture. This equipment can sometimes be represented
by a uniform load (known as a collateral load), but
the points of attachment are usually subjected to con-
centrated loads, which require a separate analysis to
account for the localized effects.
2. Live load: This load represents the force imposed on
the structure by the occupancy and use of the build-
ing. Building codes give minimum design live loads
in pounds per square foot, which vary with the clas-
sification of occupancy and use. While live loads are
expressed as uniform, as a practical matter any occu-
pancy loading is inevitably nonuniform. The degree
of nonuniformity that is acceptable is a matter of
engineering judgment. Some building codes deal with
nonuniformity of loading by specifying concentrated
loads in addition to uniform loading for some occu-
pancies. In an industrial building, often the use of the
building may require a live load in excess of the code-
stated minimum. Often this value is specified by the
owner or calculated by the engineer. Also, the loading
may be in the form of significant concentrated loads
as in the case of storage racks or machinery.
3. Snow load: Most codes differentiate between roof live
load and snow load. Snow loads are a function of local
climate, roof slope, roofing type, terrain, building
internal temperature, and building geometry. These
factors may be treated differently by various codes.
4. Rain load: This load is now recognized as a separate
loading condition. In the past, rain was accounted for
in the live load. However, some codes have a more
refined standard. Rain loading can be a function of
storm intensity, roof slope, and roof drainage. There is
also the potential for rain on snow in certain regions.
5. Wind load: This load is well codified and a function
of local climate conditions, building height, building
geometry, and exposure as determined by the sur-
rounding environment and terrain. Building codes
account for increases in local pressure at edges and
corners and often have stricter standards for indi-
vidual components than for the gross building. Wind
can apply both inward and outward forces to various
surfaces on the building exterior and can be affected
by the size of wall openings. Where wind forces pro-
duce overturning or net upward forces, there must be
an adequate counterbalancing structural dead weight,
or the structure must be anchored to an adequate
foundation.
6. Earthquake load: Seismic loads are established by
building codes and are based on:
a. The degree of seismic risk.
b. The degree of potential damage.
c. The possibility of total collapse.
d. The feasibility of meeting a given level of
protection.
Earthquake loads in building codes are usually equiv-
alent static loads. Seismic loads are generally a func-
tion of:
a. The geographical and geological location of the
building.
b. The use of the building.
c. The nature of the building structural system.
d. The dynamic properties of the building.
e. The dynamic properties of the site.
f. The weight of the building and the distribution of
the weight.
Load combinations are formed by adding the effects of
loads from each of the load sources cited in the preceding

text. Codes or industry standards often give specific load
combinations that must be satisfied. It is not always neces-
sary to consider all loads at full intensity. Also, certain loads
are not required to be combined at all. For example, wind
loads need not be combined with seismic loads. In some
cases, only a portion of a load must be combined with other
loads. When a combination does not include loads at full
intensity, it represents a judgment as to the probability of
simultaneous occurrence with regard to time and intensity.

Chapter 2
Owner-Established Criteria
Every industrial building is unique. Each is planned and con-
structed to requirements relating to building usage, the pro-
cess involved, specific owner requirements and preferences,
site constraints, cost, and building regulations. The process
of design must balance all of these factors. The owner must
play an active role in communicating to the designer all
requirements specific to the building such as:
1. Area, bay size, plan layout, aisle location, future
expansion provisions
2. Clear heights
3. Relationship between functional areas, production
flow, acoustical considerations
4. Exterior appearance
5. Materials and finishes
6. Machinery, equipment, and storage method
7. Loads
There are instances where loads in excess of code mini-
mums are required. Such cases call for owner involvement.
The establishment of loading conditions provides a struc-
ture of adequate strength. A related set of criteria are needed
to establish the serviceability behavior of the structure.
Serviceability design considers such topics as deflection,
drift, vibration, and the relation of the primary and second-
ary structural systems and elements to the performance of
nonstructural components such as roofing, cladding, and
equipment. Serviceability issues are not strength issues, but
rather maintenance and human response considerations. Ser-
viceability criteria are discussed in detail in AISC Design
Guide 3, Serviceability Design Considerations for Steel
Buildings (West et al., 2003), hereafter referred to as AISC
Design Guide 3. Criteria taken from the Design Guide are
presented in this text as appropriate.
As can be seen from this discussion, the design of an indus-
trial building requires active owner involvement. This is also
illustrated by the following topics: slab-on-ground design,
jib cranes, interior vehicular traffic, and future expansion.
2.1 SLABS-ON-GROUND DESIGN
One important aspect to be determined in an industrial build-
ing design is the specific loading to which the floor slab will
be subjected. Forklift trucks, rack storage systems, or wood
dunnage supporting heavy manufactured items cause con-
centrated loads in industrial structures. The important point
here is that these loadings are nonuniform. The slab-on-
ground is thus often designed as a plate on an elastic founda-
tion subject to concentrated loads.
It is common for owners to specify that slabs-on-ground
be designed for a specific uniform loading (e.g., 500 psf). If a
slab-on-ground is subjected to a uniform load, it will develop
no bending moments. Minimum thickness and no reinforce-
ment would be required. Real loads are not uniform, and an
analysis using an assumed nonuniform load or the specific
concentrated loading for the slab is required. An excel-
lent reference for the design of slabs-on-ground is Design-
ing Floor Slabs on Grade (Ringo and Anderson, 1996). In
addition, the following guides provide useful information:
the ACI Guide for Concrete Floor and Slab Construction,
ACI 302.1R-15 (ACI, 2015), and Guide to Design of Slabs-
on-Ground, ACI 360R-10 (ACI Committee 360, 2010).
2.2 JIB CRANES
Another loading condition that should be considered is the
installation of jib cranes. Often the owner has plans to install
such cranes at some future date, but because they are a pur-
chased item and often installed by plant engineering person-
nel or the crane manufacturer, the owner may inadvertently
neglect them during the design phase.
Jib cranes have a horizontal member known as a jib or
boom that supports a moveable hoist fixed to a wall or col-
umn. Jib cranes that are simply added to a structure can cre-
ate a myriad of problems, including column distortion and
misalignment, column bending failures, crane runway and
crane rail misalignment, and excessive column base shear.
It is essential to know the location and size of jib cranes
in advance so that columns can be properly designed and
proper bracing can be installed if needed. Columns support-
ing jib cranes should be designed to limit the deflection at
the end of the jib boom to the boom length divided by 225.
2.3 INTERIOR VEHICULAR TRAFFIC
The designer must establish the exact usage to which the
structure will be subjected. Interior vehicular traffic is a
major source of problems in structures. Forklift trucks can
accidentally buckle the flanges of a column, shear off anchor
rods in column bases, and damage walls.
Proper consideration and handling of the forklift truck
problem may include some or all of the following:
1. Use of masonry or concrete exterior walls in lieu
of metal panels. (Often the lowest section of wall

is masonry or concrete, and metal panels are used
above.)
2. Installation of fender posts (bollards) for columns and
walls may be required where speed and size of fork
trucks are such that a column or load bearing wall
could be severely damaged or collapsed upon impact.
3. Use of metal guard rails or steel plate adjacent to wall
elements may be in order.
4. Curbs.
Lines defining traffic lanes painted on factory floors have
never been successful in preventing structural damage from
interior vehicular operations. The only realistic approach for
solving this problem is to anticipate potential impact and
damage and to install barriers and/or materials that can with-
stand such abuse.
2.4 FUTURE EXPANSION
Except where no additional land is available, every industrial
structure is a candidate for future expansion. Lack of plan-
ning for such expansion can result in considerable expense.
When consideration is given to future expansion, there are
a number of practical considerations that require evaluation.
1. The direction of principal and secondary framing
members requires study. In some cases, it may prove
economical to have a principal frame line along a
building edge where expansion is anticipated and to
design edge beams, columns, and foundations for the
future loads. If the structure is large and any future
expansion would require creation of an expansion
joint at a juncture of existing and future construction,
it may be prudent to have that edge of the building
consist of nonload-bearing elements. Foundation
design must also include provision for expansion.
2. Roof drainage must be considered. An addition that
is constructed with low points at the junction of the
roofs can present serious problems in terms of water,
ice, and snow piling effects.
3. Lateral stability to resist wind and seismic loads
is often provided by X-bracing in walls or by shear
walls. Future expansion may require removal of such
bracing. The structural drawings should indicate the
critical nature of wall bracing and its location to pre-
vent accidental removal. In this context, bracing can
interfere with many plant production activities, and
the importance of such bracing cannot be overempha-
sized to the owner and plant engineering personnel.
Bracing should be located to provide the capability
for future expansion without its removal.
2.5 DUST CONTROL/EASE OF MAINTENANCE
In certain buildings (e.g., food processing plants), dust
control is essential. Ideally there should be no horizontal
surfaces on which dust can accumulate. Hollow structural
section (HSS) purlins reduce the number of horizontal sur-
faces compared to joists or C- or Z-shaped sections. If hori-
zontal surfaces can be tolerated in conjunction with a regular
cleaning program, C- or Z-shaped sections may be prefer-
able to joists. The same thinking should be applied to the
selection of main framing members (i.e., HSS or box sec-
tions may be preferable to wide-flange sections or trusses).
2.6 ELECTRICAL, PIPING, AND EQUIPMENT
LOADS
The owner must indicate loads and locations of electri-
cal, piping, and equipment loads. Process piping should
be assumed to be full when calculating the loads on the
structural system. Ductwork can be very critical to the load
effects on the structure. It is wise to consider ductwork to be
a minimum of half full and to consider the wet density of the
material in the duct. Depending on the support system for
equipment, temperature effects should be investigated.
The designer must also be aware of special concentrated
loads as dictated in the 2015 International Building Code
(ICC, 2015), hereafter referred to as the IBC, Section 1607.4
and Table 1607.1.
Snow drifts can result from rooftop equipment and piping.
Minimum Design Loads and Associated Criteria for Build-
ings and Other Structures (ASCE, 2016), hereafter referred
to as ASCE/SEI 7-16, should be consulted for the calcula-
tion of snow drifting.

Chapter 3
Roof Systems
The roof system is often the most expensive part of an indus-
trial building (even though walls are costlier per square foot).
Designing for a 20-psf mechanical surcharge load when only
10 psf is required adds cost over a large area.
Often the premise that guides the design is that the owner
will always be hanging new piping or installing additional
equipment, and a prudent designer will allow for this in the
system. If this practice is followed, the owner should be con-
sulted, and the decision to provide excess capacity should
be that of the owner. The design live loads and collateral
(equipment) loads should be clearly identified on the struc-
tural plans.
3.1 STEEL DECK FOR BUILT-UP OR
MEMBRANE ROOFS
Decks are commonly 12 in. deep, but deeper units are also
available. The Steel Deck Institute (SDI) Standard for Steel
Roof Deck (SDI, 2017b), has identified three standard pro-
files for 12-in. steel deck—narrow rib, intermediate rib, and
wide rib—and has published load tables for each profile for
thicknesses varying from 0.0299 in. to 0.0478 in. (nominally
22 gage to 16 gage). These three profiles, listed in Table 3-1,
are identified as NR, IR, and WR, and correspond to the
manufacturers’ designations A, F, and B, respectively. SDI
identifies the standard profile for 3-in. deck as 3DR. A com-
parison of weights for each profile in various gages shows
that strength-to-weight ratio is most favorable for wide-rib
deck and least favorable for narrow-rib deck. In general, the
deck selection that results in the least weight per square foot
may be the most economical. However, consideration must
also be given to the flute width because the insulation must
span the flutes. In the northern areas of the United States, high
roof loads and thick insulation generally make the wide-rib
profile predominant. In the South, low roof loads and thin-
ner insulation make the intermediate-rib profile common.
Where very thin insulation is used, narrow-rib deck may be
required, although this is not a common profile. In general,
the lightest weight deck consistent with insulation thickness
and span should be used.
In addition to the load, span, and thickness relations estab-
lished by the load tables, there are other considerations in
the selection of a profile and gage for a given load and span.
First, SDI limits deflection due to a 200-lb concentrated
load at midspan to the span divided by 240. Secondly, the
SDI Code of Standard Practice (SDI, 2017a) has published
a table of maximum recommended spans for construc-
tion and maintenance loads, and, finally FM Global’s Loss
Prevention Data for Roofing Contractors (FM Global, 2019)
lists maximum spans for various profiles and gages, which
are shown in Table 3-2.
FM Global requires a sidelap fastener between supports.
This fastener prevents adjacent panels from deflecting dif-
ferentially when a load exists at the edge of one panel but
not on the edge of the adjacent panel. FM Global permits
an overspan from its published tables of 6 in. (previously
an overspan of 10% had been allowed) when “necessary to
accommodate column spacing in some bays of the building.
It should not be considered an original design parameter.”
SDI recommends that the sidelaps in cantilevers be fastened
at 12 in. on center.
Steel deck can be attached to supports by welds or fas-
teners, which can be power or pneumatically installed or
self-drilling and self-tapping. The SDI Standard for Steel
Roof Deck requires a maximum attachment spacing of 18 in.
along supports. FM Global requires the use of 12-in. spacing
as a maximum, which is more common. The attachment of
roof deck must be sufficient to provide bracing to the struc-
tural roof members, to anchor the roof to prevent uplift, and,
in many cases, to serve as a diaphragm to carry lateral loads
to the bracing. While the standard attachment spacing may
be acceptable in many cases, decks designed as diaphragms
may require additional connections. Diaphragm capacities
can be determined from the SDI Diaphragm Design Manual
(SDI, 2015).
Manufacturers of metal deck are constantly researching
ways to improve section properties with maximum econ-
omy. Considerable differences in cost may exist between
prices from two suppliers of identical deck shapes; there-
fore, the designer is urged to research the cost of the deck
system carefully. A few cents per square foot savings on a
large roof area can mean a significant savings to the owner.
Several manufacturers can provide steel roof deck and
wall panels with special acoustical surface treatments for
specific building use. Properties of such products can be
obtained from the manufacturers. Special treatment for
acoustical reasons must be specified by the owner.
3.2 METAL ROOFS
Standing-seam roof systems were first introduced in the late
1960s, and today many manufacturers produce standing
seam panels. A difference between the standing-seam roof
and lap-seam roof (through-fastener roof) is the manner in
which two panels are joined to each other. The seam between
two standing-seam panels is made in the field with a tool that

Table 3-1. SDI Recommended Spans*
Recommended Maximum Spans for Construction
and Maintenance Loads for Standard 1½-in. and 3-in. Roof Deck
Deck Type
Span
Condition
Gage
Number
ASD Span
(ft-in.)
ASD Cantilever
Span (ft-in.)
NARROW
RIB
NR22
Single
22 2'-11" 0'-10"
NR20 20 3'-08" 1'-00"
NR18 18 5'-00" 1'-03"
NR16 16 6'-05" 1'-07"
NR22
Double
or
Triple
22 3'-07"
NR20 20 4'-06"
NR18 18 6'-02"
NR16 16 7'-11"
INTERMEDIATE
RIB
IR22
Single
22 3'-05" 0'-11"
IR20 20 4'-03" 1'-01"
IR18 18 5'-10" 1'-06"
IR16 16 7'-06" 1'-10"
IR22
Double
or
Triple
22 4'-03"
IR20 20 5'-03"
IR18 18 7'-02"
IR16 16 9'-03"
WIDE
RIB
WR22
Single
22 5'-08" 1'-06"
WR20 20 7'-00" 1'-10"
WR18 18 9'-06" 2'-05"
WR16 16 12'-02" 3'-00"
WR22
Double
or
Triple
22 6'-11"
WR20 20 8'-07"
WR18 18 11'-08"
WR16 16 15'-00"
DEEP
RIB
DR22
Single
22 11'-11" 3'-04"
DR20 20 15'-04" 4'-02"
DR18 18 21'-01" 5'-07"
DR16 16 27'-05" 7'-01"
DR22
Double
or
Triple
22 14'-07"
DR20 20 18'-11"
DR18 18 26'-00"
DR16 16 33'-09"
* From the Manual of Construction with Steel Deck (SDI, 2016). Reproduced with permission of the Steel Deck Institute.
Spans shown are calculated using 33-ksi steel and allowable strength design (ASD) and are considered to be conservative. Longer spans may be permitted
when using load and resistance factor (LRFD) designs or for higher strength steels. Consult the deck manufacturer for further guidance.

makes a cold-formed, weather-tight joint. (Note: Some pan-
els can be seamed without special tools.) The joint is made at
the top of the panel. The standing-seam roof is also unique in
the manner in which it is attached to the purlins. The attach-
ment is made with a clip concealed inside the seam. This clip
secures the panel to the purlin and may allow the panel to
move when experiencing thermal expansion or contraction.
In standing-seam panels, a continuous single-skin mem-
brane results after the sidelap seam is made because through-
the-roof fasteners have been eliminated. The elevated seam
and single-skin member provides a watertight system. The
ability of the roof to experience unrestrained thermal move-
ment eliminates damage to insulation and structure caused
by temperature effects that built-up and through-fastened
roofs commonly experience. Thermal spacer blocks are
often placed between the panels and purlins in order to
ensure a consistent thermal barrier. Due to the superiority of
the standing-seam roof, most manufacturers are willing to
offer considerably longer guarantees than those offered on
lap-seam roofs.
Because of the ability of standing-seam roofs to move on
sliding clips, they possess only minimal diaphragm strength
and stiffness. The designer should assume that the standing-
seam roof has no diaphragm capacity and, in the case of steel
joists, should specify that sufficient bridging be provided to
laterally brace the joists under design loads.
The reader is referred to A Design Guide for Standing
Seam Roof Panels, AISI CF00-1 (AISI, 2000), for further
information on standing-seam roofs.
3.3 INSULATION AND ROOFING
Duetoenergyconcerns,theuseofadditionaland/orimproved
roof insulation has become common. Coordination with the
mechanical requirements of the building is necessary. Gen-
erally, the use of additional insulation is warranted, but there
are at least two practical problems that occur as a result.
Less heat loss through the roof results in greater snow and
ice build-up and larger snow loads. As a consequence of the
same effect, the roofing is subjected to colder temperatures
and, for some systems (built-up roofs), thermal movement,
which may result in cracking of the roofing membrane.
3.4 EXPANSION JOINTS
Althoughindustrialbuildingsareoftenconstructedofflexible
materials, roof and structural expansion joints are required
when horizontal dimensions are large. It is not possible to
state exact requirements relative to distances between expan-
sion joints because of the many variables involved, such as
ambient temperature during construction and the expected
temperature range during the life of the structure. An excel-
lent reference on the topic of thermal expansion in buildings
and location of expansion joints is the Federal Construction
Council’s Expansion Joints in Buildings (Federal Construc-
tion Council, 1974). The report presents the figure shown
here as Figure 3-1 as a guide for spacing structural expan-
sion joints in beam and column frame buildings based on
design temperature change. The report includes data for
numerous cities and gives modifying factors that are applied
to the allowable building length as appropriate.
The report indicates that the curve is directly applicable
to buildings of beam-and-column construction hinged at the
base with heated interiors. When other conditions prevail,
the following rules are applicable:
1. If the building will be heated only and will have
hinged-column bases, use the allowable length as
specified.
2. If the building will be air conditioned as well as
heated, increase the allowable length by 15% (if the
environmental control system will run continuously).
3. If the building will be unheated, decrease the allow-
able length by 33%.
4. If the building will have fixed-column bases, decrease
the allowable length by 15%.
5. If the building will have substantially greater stiffness
against lateral displacement in one direction, decrease
the allowable length by 25%.
Table 3-2. FM Global Data
Types 1.5A, 1.5F, 1.5B, and 1.5BI Deck
Nominal 1½-in. Depth/No Stiffening Grooves
22 Gage 20 Gage 18 Gage
Type 1.5A
(Narrow rib)
4'-10" 5'-3" 6'-0"
Type 1.5F
(Intermediate rib)
4'-11" 5'-5" 6'-3"
Type 1.5B, Bl
(Wide rib)
6'-0" 6'-6" 7'-5"

When more than one of these design conditions prevails
in a building, the percentile factor to be applied should be
the algebraic sum of the adjustment factors of all the various
applicable conditions.
Regarding the type of structural expansion joint, most
engineers agree that the best method is to use a line of dou-
ble columns to provide a complete separation at the joints.
When joints other than the double column type are employed,
low-sliding elements, such as shown in Figures 3-2 and 3-3,
are generally used. Slip connections may induce some level
of inherent restraint to movement due to binding or debris
build-up.
Very often buildings may be required to have fire walls
in specific locations. Fire walls may be required to extend
Fig. 3-1. Expansion joint spacing graph (Federal Construction Council, 1974).
Fig. 3-2. Beam expansion joint.

above the roof or they may be allowed to terminate at the
underside of the roof. Such fire walls become locations for
expansion joints. In such cases the detailing of joints can be
difficult. Figures 3-2 and 3-3 depict typical details to permit
limited expansion.
Expansion joints in the structure should always be car-
ried through the roofing. Additionally, depending on mem-
brane type, other joints called area dividers are necessary
in the roof membrane. These joints are membrane relief
joints only and do not penetrate the roof deck. Area divider
joints are generally placed at intervals of 150 to 250 ft for
adhered membranes, at somewhat greater intervals for bal-
lasted membranes, and 100 to 200 ft in the case of steel
roofs. Spacing of joints should be verified with manufac-
turer’s requirements. The range of movement between joints
is limited by the flexibility and movement potential of the
anchorage scheme and, in the case of standing-seam roofs,
by the clip design. Manufacturer’s recommendations should
be consulted and followed. Area dividers can also be used
to divide complex roofs into simple squares and rectangles.
3.5 ROOF PITCH, DRAINAGE, AND PONDING
Prior to determining a framing scheme and the direction of
primary and secondary framing members, it is important
to decide how roof drainage is to be accomplished. If the
structure is heated, interior roof drains may be justified. For
unheated spaces, exterior drains and gutters may provide the
solution.
For some building sites, it may not be necessary to have
gutters and downspouts to control storm water, but their use
is generally recommended or required by the owner. Signifi-
cant operational and hazardous problems can occur where
Fig. 3-3. Joist expansion joint.

unless the roof surface is configured to prevent the accumu-
lation of water.” The possibility of plugged drains means that
the load at the initiation of ponding must include the depth
of impounded water at the elevation of overflow drains, roof
edges, or scuppers. It is clear from reading the AISC Specifi-
cation that it is not necessary to include the weight of water
that would accumulate after the “initiation of ponding.”
Where snow load is used by the code, the designer may be
required to add 5 psf to the roof load to account for the effect
of rain on snow. Also, consideration must be given to areas
of drifted snow. It is clear that judgment must be used in the
determination of loading at the initiation of ponding. It is
equally clear that 100% of the roof design load would rarely
be appropriate for the loading at the initiation of ponding.
A continuously framed or cantilever system may be more
critical than a simple-span system. With continuous framing,
rotations at points of support due to nonuniformly distributed
roof loads will initiate upward and downward deflections in
alternate spans. The water in the uplifted bays drains into
the adjacent downward deflected bays, compounding the
effect and causing the downward deflected bays to approach
the deflected shape of simple spans. For these systems, one
approach to ponding analysis could be based on simple
beam stiffness, although a more refined analysis could be
used. The designer should also consult with the plumbing
designer to establish whether or not a controlled flow (water
retention) drain scheme is being used. Such an approach
allows the selection of smaller pipes because the water is
impounded on the roof and slowly drained away.
A situation that is not addressed by building code drainage
design is shown in Figure 3-4. The author has investigated
several roof ponding collapses where the accumulation of
water is greater than would be predicted by drainage analysis
for the area shown in Figure 3-4. As the water drains toward
the eave, it finds the least resistance to flow along the parapet
to the aperture of the roof. Designers are encouraged to pay
close attention to these types of situations and to provide a
conservative design for ponding in the aperture area.
Besides rainwater accumulation, the designer should give
consideration to excessive build-up of material on roof sur-
faces from industrial operations, such as fly ash and other
airborne material. Enclosed valleys, parallel high- and low-
aisle roofs, and normal wind flows can cause unexpected
build-ups and possibly roof overload.
3.6 JOISTS AND PURLINS
A decision must be made whether to span the long direction
of bays with the main beams, trusses, or joist girders that
support short-span joists or purlins or to span the short direc-
tion of bays with main framing members that support longer
span joists or purlins. Experience in this regard is that span-
ning the shorter bay dimension with primary members will
water is discharged at the eaves or scuppers in cold climates,
causing icing of ground surfaces and hanging of ice from the
roof edge. This is particularly a problem at overhead door
locations and may occur with or without gutters. Protec-
tion from falling ice must be provided at all building service
entries.
Performance of roofs with positive drainage is gener-
ally good. Due to problems that result from poor drainage,
such as ponding, roofing deterioration, and leaking, the IBC
requires a roof slope of at least 4 in. per ft.
Ponding, which is often not understood or is overlooked,
is a phenomenon that may lead to severe distress or partial
or general collapse. As it applies to roof design, ponding has
two meanings. To the roofing industry, ponding describes the
condition in which water accumulated in low spots has not
dissipated within 24 hours of the last rain storm. Ponding of
this nature is addressed in roof design by positive roof drain-
age and control of the deflections of roof framing members.
As a structural engineering issue, ponding is a load/deflec-
tion situation, in which there is incremental accumulation of
rain water in the deflecting structure. The purpose of a pond-
ing check is to ensure that equilibrium is reached between
the incremental loading and the incremental deflection. This
convergence must occur at a level of stress that is within the
available value.
AISC Specification for Structural Steel Buildings, ANSI/
AISC 360-16 (AISC, 2016b), hereafter referred to as the
AISC Specification, gives procedures in Appendix 2 for
addressing the problem of ponding where roof slopes and
drains may be inadequate. The simplified design for pond-
ing method is expressed in AISC Specification Equations
A-2-1 through A-2-4. These relations control the stiffness of
the primary- and secondary-framing members and the deck.
This method, however, can produce unnecessarily conserva-
tive results.
A more exact method is provided in AISC Specification
Appendix 2, Section 2.2, Improved Design for Ponding. The
key to the use of the improved method is the calculation of
stress in the framing members due to loads present at the
initiation of ponding. The difference between 0.8Fy and the
initial stress is used to establish the required stiffness of the
roof framing members. The initial stress (“at the initiation of
ponding”) is determined from the loads present at that time.
These should include all or most of the dead load and may
include some portion of snow, rain, or live load.
Steel Joist Institute (SJI) Structural Design of Steel Joist
Roofs to Resist Ponding Loads, Technical Digest #3 (SJI,
2007), provides additional information on ponding.
The amount of accumulated water used in the design for
ponding is also subject to judgment. AISC Specification
Section B3.10, Design for Ponding, states that, “The roof
system shall be investigated through structural analysis to
ensure strength and stability under ponding conditions,

T
Fig. 3-4. Aperture drainage.
provide the most economical system. However, this decision
may not be based solely on economics but rather on such
factors as ease of erection, future expansion, direction of
crane runs, or location of overhead doors.
On the use of steel joists or purlins, experience again
shows that each case must be studied. SJI Standard Specifi-
cations, Load Tables, and Weight Tables for Steel Joists and
Joist Girders (SJI, 2015b), hereafter referred to as the SJI
Specification, are based upon distributed loads only. Modifi-
cations for concentrated loads should be made in accordance
with the SJI Code of Standard Practice for Steel Joists and
Joist Girders (SJI, 2015a). Significant concentrated loads
should be supported by hot-rolled framing members. How-
ever, in the absence of large concentrated loads, joist framing
can generally be more economical than hot-rolled framing.
Cold-formed C- and Z-shaped purlins provide an alterna-
tive to rolled wide-flange sections. The provisions contained
in the North American Specification for the Design of Cold-
Formed Steel Structural Members, AISI S100-16 (AISI,
2016a), hereafter referred to as theAISI Specification, should
be used for the design of cold-formed steel purlins. The AISI
Design Guide for Cold-Formed Steel Purlin Roof Framing
Systems (AISI, 2009) also provides design examples for the
design of cold-formed purlins. Additional economy can be
achieved with C- and Z-shaped sections because they can be
designed and constructed as continuous members. However,
progressive failure should be considered if there is a possi-
bility for a loss in continuity after installation.
Other considerations in the use of C- and Z-shaped sec-
tions include:
1. Z-shaped sections ship economically due to the fact
that they can be “nested.”
2. Z-shaped sections can be loaded through the shear
center; C-shaped sections cannot.
3. On roofs with appropriate slope, a Z-shaped section
will have one principal vertical axis, while a C-shaped
section provides this condition only for flat roofs.
4. Many erectors indicate that lap-bolted connections for
C- or Z-shaped sections are more expensive than the
simple, welded-down connections for joist ends.
5. At approximately a 30-ft span length, C- and Z-shaped
sections may cost about the same as a joist for the
same load per foot. For shorter spans, C- and Z-shaped
sections are normally less expensive than joists.
3.7 ROOF PENETRATIONS AND EQUIPMENT
When headers are used to support rooftop equipment, the
maximum size of an opening is one that can fit between two
beams or joists without disrupting the specified beam spac-
ing for a given framing situation. Openings often coincide
with additional concentrated loads, such as at rooftop units
or other types of equipment. Curbs can be set atop the steel
roof deck and may be screwed directly to the deck. The deck
opening is cut to match the inside dimensions of the curb.
Headers or a small frame should be provided to carry the
curb load to the joists. Wood or steel blocking is often placed
between the deck flutes to prevent the deck from crushing
between the curb and the headers. A typical header detail is
shown in Figure 3-5.
When joist framing is used, it is always desirable to locate
the concentrated loads on panel points and thus eliminate
top-chord bending. Small isolated openings for vents can
usually be shifted to align with panel points. This, how-
ever, requires that the opening frame is made to conform
to the panel-point spacing. For repetitive openings with a
consistent pattern, special joists designed for uniform and
concentrated loads can be used. The frames are usually con-
structed from hot-rolled angles that have been welded into
the required shapes. The vertical leg of the header angle is
coped or a short piece of angle is welded to the end of the
header to create a seat.

Fig. 3-5. Header conforming to panel point spacing.

Chapter 4
Roof Trusses
Primary roof framing for a conventionally designed indus-
trial building generally consists of wide-flange beams, steel
joist girders, or fabricated trusses. For relatively short spans
of 30 to 40 ft, steel beams provide an economical solution,
particularly if a multitude of hanging loads is present. For
spans greater than 30 ft, steel joist girders are often used
to support roof loads. Fabricated steel roof trusses are often
used for spans greater than 80 ft. In recent years, little has
been written about the design of steel roof trusses. Most text-
books addressing the design of trusses were written when
riveted connections were used. Today, welded trusses and
field-bolted trusses are used exclusively. Presented in the
following paragraphs are concepts and principles that apply
to the design of roof trusses.
4.1 GENERAL DESIGN AND ECONOMIC
CONSIDERATIONS
No absolute statements can be made about what truss con-
figuration will provide the most economical solution for a
particular situation; however, the following statements can
be made regarding truss design.
1. Span-to-depth ratios of 15 to 20 generally prove to
be economical; however, shipping depth limitations
should be considered so that shop fabrication can be
maximized. The maximum depth for shipping is con-
servatively 14 ft. Greater depths will require the web
members to be field bolted or field welded, which may
increase erection costs.
2. The length between splice points is also limited by
shipping lengths. The maximum shippable length
varies according to the destination of the trusses, but
lengths of 80 ft are generally shippable, and 100 ft
is often possible. Because maximum available mill
length is approximately 70 ft, the distance between
splice points is typically set at a maximum of 70 ft.
Greater distances between splice points will generally
require truss chords to be shop spliced.
3. In general, the rule “deeper is cheaper” is true; how-
ever, the costs of additional lateral bracing for more
flexible truss chords must be carefully examined rela-
tive to the cost of larger chords that may require less
lateral bracing. The lateral bracing requirements for
the top and bottom chords should be considered inter-
actively while selecting chord sizes and types. Par-
ticular attention should be paid to loads that produce
compression in the bottom chord. In this condition,
additional chord bracing will most likely be necessary.
4. If possible, truss depths should be selected so that tees
can be used for the chords rather than wide-flange
shapes. Tees can eliminate or reduce the need for gus-
set plates.
5. Higher strength steels (Fy > 50 ksi) usually result in
more efficient truss members.
6. Web arrangements are illustrated in Figures 4-1 and
4-2, which generally provide economical web systems.
7. Only a few web angle sizes should be selected, and
efficient long-leg angles should be utilized for greater
resistance to buckling. Differences in angle sizes
should be recognizable. For instance, avoid using an
L4×3×4 and an L4×3×c in the same truss.
8. HSS or pipe sections may prove to be more effective
web members at some web locations; however, they
may increase fabrication cost due to increased fit-up
time and welding.
9. Designs using the LRFD load combinations from
ASCE/SEI 7-16 will often lead to truss savings when
heavy, long-span trusses are required. This is due to
the higher DL-to-LL ratios for these trusses.
10. The weight of gusset plates, shim plates, and bolts can
be significant in large trusses. This weight must be
considered in the design because it often approaches
10 to 15% of the truss weight.
11. In computer analyses of trusses where rigid joints are
assumed, secondary bending moments will show up
in the analysis. The reader is referred to Nair (1988a)
wherein it is suggested that as long as these second-
ary stresses do not exceed 4,000 psi, they may be
neglected. Secondary stresses should not be neglected
if the beneficial effects of continuity are being con-
sidered in the design process—for example, effective
length determination. The designer must be consis-
tent. That is, if the joints are considered as pins for
the determination of forces, then they should also be
considered as pins in the design process. The assump-
tion of rigid joints in some cases may provide uncon-
servative estimates on the deflection of the truss.
12. Repetition is beneficial and economical. Use as few
different truss depths as possible. It is cheaper to vary
the chord size rather than the truss depth.
13. Wide-flange chords with gussets may be necessary
when significant bending moments exist in the chords
(i.e., subsystems not supported at webs or large dis-
tances between webs).

14. Design and detailing of long-span joists and joist gird-
ers should be in accordance with the SJI Specification.
4.2 CONNECTION CONSIDERATIONS
The following are some issues to consider relative to the
various types of connections involved in truss design.
1. Tee chords are generally economical because they can
eliminate gusset plates. The designer should exam-
ine the connection requirements to determine if the
tee stem is, in fact, long enough to eliminate gusset
requirements. The use of a deeper tee stem is gener-
ally more economical than adding numerous gusset
plates even if this means an addition in overall weight.
Adding tee stems will usually require complete-joint-
penetration (CJP) welds between the gusset plate
and the tee stem, which may increase fabrication and
inspection costs.
2. Block shear requirements and the effective area in
compression should be carefully checked in tee stems
and gussets. Shear rupture of chord members at panel
points should also be investigated because this can
often control wide-flange chords.
3. Intermediate connectors such as stitch fasteners or
fillers may be required for double-web members.
4. If wide-flange chords are used with wide-flange web
members, it is generally more economical to orient the
chords with their webs horizontal. Gusset plates for
the web members can then be either bolted or welded
to the chord flanges. To eliminate the cost of fabricat-
ing large shim or filler plates for the diagonals, the use
of comparable depth wide-flange diagonals should be
considered.
5. When trusses require field-bolted joints, the use of
slip-critical bolts in conjunction with oversize holes
will allow for erection alignment. Also, if standard
holes are used with slip-critical bolts and field fit-up
problems occur, holes can be reamed without signifi-
cantly reducing the allowable bolt shears.
6. For the end connection of trusses, top-chord seat-type
connections should also be considered. Seat connec-
tions allow more flexibility in correcting column-truss
alignment during erection. Seats also provide for effi-
cient erection and are more stable during erection than
bottom bearing trusses. When seats are used, a simple
bottom-chord connection is recommended to prevent
the truss from rolling during erection.
7. For symmetrical trusses, a center splice should be
used to simplify fabrication, even though forces may
be larger than for an offset splice.
8. End plates can provide efficient compression splices.
9. It is often less expensive to locate the work point of
the end diagonal at the face of the supporting member
rather than designing the connection for the eccen-
tricity between the column centerline and the face of
the column. When this is done, the column should be
designed for the eccentricity of load.
4.3 TRUSS BRACING
Stability bracing is required at discrete locations where the
designer requires bracing for the design of the members in
a truss. These locations are generally at panel points. Brac-
ing requirements are provided in AISC Specification Appen-
dix 6. To function properly, braces must have sufficient
strength and stiffness.As a general rule, the stiffness require-
ment will control the design of the bracing unless the stiff-
ness is derived from axial stresses only. Braces that displace
due to axial loads are very stiff, thus strength requirements
generally control.
Fig. 4-1. Economical truss web arrangement.
Fig. 4-2. Economical truss web arrangement.

The Association for Iron and Steel Technology (AIST)
Guide for the Design and Construction of Mill Buildings
(AIST, 2003), hereafter referred to as AIST TR-13, requires
a 0.025P force for bracing. AIST TR-13 is silent on stiffness
requirements.
Designers must determine the number of “out-of-straight”
trusses that should be considered for a given bracing situ-
ation. No definitive rules exist; however, the Australian
Code (BCA, 2015) indicates that no more than seven out-of-
straight members need to be considered. For columns, Chen
and Tong (1994) recommend that n columns be considered
in the out-of-straight condition, where n is the total number
of columns in a story. This suggests that n trusses could be
considered in the bracing design. The number to be consid-
ered is rounded up to a whole number. Thus, if 10 trusses
were to be braced, bracing forces would be based on four
trusses. Common practice is to provide horizontal bracing
every five to six bays to transfer bracing forces to the main
force-resisting system.
In addition to stability bracing, top- and bottom-chord
bracing may also be required to transfer lateral loads to the
main lateral-stability system. The force requirements for the
lateral loads must be added to the stability force require-
ments. Lateral load bracing is placed in either the plane of
the top chord or the plane of the bottom chord, but generally
not in both planes.
Requirements for truss bottom-chord bracing are also
discussed in “The Importance of Tension Chord Bracing”
(Fisher, 1983).
4.3.1 Roof Truss Stability Bracing Example
For the truss system shown in Figure 4-3, determine the
brace forces in the web members (tension-only X-bracing)
of the horizontal truss. Use the requirements from AISC
Specification Appendix 6. For illustration purposes the
Fig. 4-3. Horizontal bracing system.

forces shown in Figure 4-3 can be considered LRFD or ASD
forces. The compression forces in the top chord in each truss
are shown in the truss elevation. The horizontal truss braces
six trusses, all with the same chord compressive forces.
The grid lines form a square pattern. The solution does not
reduce the number of trusses to be braced based on the Chen
and Tong (1994) paper. The axially loaded web members in
this example have adequate stiffness to satisfy AISC Specifi-
cation Appendix 6 requirements.
The horizontal truss is considered a panel bracing system;
therefore, the required shear strength of the braces for the web
members is based on AISC Specification Equation A-6-1:
V P
0.005
br r
= (Spec. Eq. A-6-1)
where
Pr =
required axial strength of the column within the
panel under consideration using LRFD or ASD load
combinations, kips
The strut forces are a function of the lateral stiffness of
the horizontal truss. If the truss has infinite stiffness, then the
strut forces would act as point bracing. See AISC Specifica-
tion Commentary Figure C-A-6.1(a). If the horizontal truss
is not rigid, the strut forces would be smaller in magnitude
than those using the point bracing equation. Conservatively
use the point brace force equation from the AISC Specifica-
tion. The AISC Specification required strength for a point
brace is:
P P
0.01
br r
= (Spec. Eq. A-6-3)
The forces in the braces do not accumulate along the
length of the truss, i.e., from grid line to grid line (Nair,
1988b). Any unbalanced shear between panels is resisted by
lateral shears in the top chord of the horizontal truss. The
forces in the bracing struts accumulate based on the number
of trusses being braced by the horizontal truss. A maximum
of three struts accumulate along each grid line to deliver the
brace forces to the horizontal truss. The panel shear forces
are additive to the horizontal truss chord axial forces. The
bracing requirements are summarized in Table 4-1.
Table 4-1. Summary of Stability Bracing Web Member Forces
Web Member Panel Shear (Panel Bracing)
Members
Panel Shear =
0.005 × Average Chord Force, kips
Panel Shear
Brace Force =
cos 45°
, kips
C1-D2, D1-C2 0.005 6 trusses
600 kips 800 kips
2
21.0
( )
+
⎛
⎝
⎜
⎞
⎠
⎟ = 29.7
C2-D3, D2-C3 0.005 6 trusses
800 kips 1
,000 kips
2
27.0
( )
+
⎛
⎝
⎜
⎞
⎠
⎟ = 38.2
C3-D4, D3-C4 0.005 6 trusses 1
,000 kips 30.0
( )
( ) = 42.4
Bracing Strut Forces (Point Bracing)
Grid Lines
Strut Force =
0.01 × Average Chord Force, kips
Total Strut Force, kips
1 7 0.01 600 kips 6.00
( ) = (3)(6.00) = 18.0
2 6 0.01 600 kips 800 kips 7.00
( )
+ = (3)(7.00) = 21.0
3 5 0.01 800 kips 1
,000 kips 9.00
( )
+ = (3)(9.00) = 27.0
4 0.01
1
,000 kips 1
,000 kips
2
10.0
+
⎛
⎝
⎜
⎞
⎠
⎟ = (3)(10.0) = 30.0
Final Top Chord Forces, kips
Grid Line Top Chord Force, kips
1 to 2 600 + 21.0 = 621
2 to 3 800 + 27.0 = 827
3 to 4 1,000 + 30.0 = 1,030

4.4 ERECTION BRACING
The engineer of record is not responsible for the design
of erection bracing unless specific contract arrangements
incorporate this responsibility into the work. However,
designers must be familiar with the Occupational Safety
and Health Administration (OSHA) erection requirements,
Safety and Health Standards for the Construction Industry,
29 CFR 1926 Part R Safety Standards for Steel Structures
(OSHA, 2010b), hereafter referred to as OSHA Subpart R.
Even though the truss designer is not responsible for the
erection bracing, the designer should consider sequence and
bracing requirements in the design of large trusses in order to
provide the most cost-effective system. Large trusses require
significant erection bracing not only to resist wind and con-
struction loads, but also to provide stability until all of the
gravity load bracing is installed. Significant cost savings can
be achieved if the required erection bracing is incorporated
into the permanent bracing system.
Erection is generally accomplished by first connecting two
trusses together with strut braces and any additional erection
braces to form a stable box system. Additional trusses are
held in place by the crane or cranes until they can be tied
off with strut braces to the already-erected stable system.
Providing the necessary components to facilitate this type
of erection sequence is essential for a cost-effective project.
Additional considerations are as follows:
1. Columns are usually erected first with the lateral brac-
ing system (see Figure 4-4). If top-chord seats are
used, the trusses can be quickly positioned on top of
the columns and braced to one another. Bottom-chord
bearing trusses require that additional stability brac-
ing be installed at the truss ends while the cranes hold
the trusses in place. This can slow down the erection
sequence.
2. Because many industrial buildings require clear
spans, systems are often designed as rigid frames. By
designing rigid frames, erection is facilitated because
the side wall columns are stabilized in the plane of
the trusses once the trusses are adequately anchored to
the columns. This scheme may require larger columns
than a braced-frame system; however, economy can
generally be recovered due to a savings in bracing and
erection time.
3. Wide-flange beams, HSS, or pipe sections should be
used to laterally brace large trusses at key locations
during erection because of greater stiffness. Steel
joists can be used; however, two notes of caution are
advised.
a. Erection bracing strut forces must be provided to
the joist manufacturer, and it must be made clear
whether joist bridging and roof deck will be in
place when the erection forces are present. Large-
angle top chords in joists may be required to con-
trol the joist slenderness ratio so that it does not
buckle while serving as the erection strut.
b. Joists are often not fabricated to exact lengths, and
long-slotted holes are generally provided in joist
Fig. 4-4. Wall bracing erection sequence.

seats. Slotted holes for bolted bracing members
should be avoided because of possible slippage.
Special coordination with the joist manufacturer
is required to eliminate the slots and to provide
a suitable joist for bracing. In addition, the joists
must be at the job site when the erector wishes to
erect the trusses.
4. Wind forces on the trusses during erection can be
considerable. Refer to ASCE/SEI 7-16 for detailed
treatment of wind forces on buildings during con-
struction. The AISC Code of Standard Practice for
Steel Buildings and Bridges (AISC, 2016a), hereafter
referred to as the AISC Code of Standard Practice,
Section 7.10.3, states that “These temporary supports
shall be sufficient to secure the bare structural steel
framing or any portion thereof against loads that are
likely to be encountered during erection, including
those due to wind and those that result from erection
operations.” The projected area of all of the truss and
other roof framing members can be significant, and
in some cases the wind forces on the unsided struc-
ture are actually larger than those after the structure is
enclosed.
5. A sway frame is typically required to plumb the
trusses during erection. These sway frames should
occur every fourth or fifth bay. An elevation view
of such a truss is shown in Figure 4-5. These frames
can be incorporated into the bottom-chord bracing
system. Sway frames are also often used to transfer
forces from one chord level to another as discussed
earlier. In these cases, the sway frames must not only
be designed for stability forces, but also the required
load transfer forces.
4.5 OTHER CONSIDERATIONS
Several other issues to consider when designing roof trusses
are listed in the following.
1. Camber: Large clear-span trusses are typically cam-
bered to accommodate dead load deflections. This is
accomplished by the fabricator by either adjusting
the length of the web members in the truss and keep-
ing the top-chord segments straight or by curving the
top chord. Tees can generally be easily curved during
assembly whereas wide-flange sections may require
cambering prior to assembly. If significant top-chord
pitch is provided and if the bottom chord is pitched,
camber may not be required. The engineer of record is
responsible for providing the fabricator with the antic-
ipated dead load deflection and special cambering
requirements. The designer must carefully consider
the truss deflection and camber adjacent to walls or
other portions of the structure where stiffness changes
cause variations in deflection. This is particularly true
at building end walls, where differential deflections
may damage continuous purlins or connections.
2. Temperature changes: Connection details that can
accommodate temperature changes are typically nec-
essary. Long-span trusses that are fabricated at one
temperature and erected at a significantly different
temperature can grow or shrink significantly.
3. Diaphragm action: Roof deck diaphragm strength
and stiffness are commonly used for strength and sta-
bility bracing for joists. The diaphragm capabilities
must be carefully evaluated if it is to be used for brac-
ing of large clear-span trusses.
For a more comprehensive treatment of erection bracing
design, the reader is referred to AISC Design Guide 3.
Fig. 4-5. Sway frame.

Chapter 5
Wall Systems
The wall system in an industrial building can be chosen for a
variety of different criteria, and the cost of the wall can vary
by as much as a factor of three. Wall systems include:
1. Field-assembled metal panels
2. Factory-assembled metal panels
3. Precast concrete panels
4. Masonry walls (partial or full height)
A particular wall system may be selected over others for one
or more specific reasons, including:
1. Cost
2. Appearance
3. Ease of erection
4. Speed of erection
5. Insulating properties
6. Fire considerations
7. Acoustical considerations
8. Ease of future expansion
9. Durability of finish
10. Maintenance/cleaning considerations
Some of these factors will be discussed in the following
sections on specific systems. Other factors are not discussed
and require evaluation on a case-by-case basis.
5.1 FIELD-ASSEMBLED PANELS
Field-assembled panels consist of an outer skin element,
insulation, and in some cases, an inner liner panel. The pan-
els vary in material thickness and are typically galvanized,
galvanized prime painted suitable for field painting, or pre-
finished galvanized. Corrugated aluminum liners are also
used. When aluminum materials are used, their compatibil-
ity with steel supports should be verified with the manufac-
turer because aluminum may cause corrosion of steel. When
an inner liner is used, some form of hat section interior sub-
girts are typically provided for stiffness. The insulation is
typically fiberglass or foam. If the inner liner sheet is used as
the vapor barrier, all joints and edges should be sealed.
Specific advantages of field-assembled wall panels
include:
1. Rapid erection of panels.
2. Good cost competition, with a large number of manu-
facturers and contractors being capable of erecting
panels.
3. Quick and easy panel replacement in the event of
panel damage.
4. Openings for doors and windows that can be created
quickly and easily.
5. The panels are lightweight so that heavy equipment
is not required for erection. Large foundations and
heavy spandrels are not required.
6. Acoustic surface treatment that can be added easily to
interior wall panels at reasonable cost.
A disadvantage of field-assembled panels in high-
humidity environments can be the formation of frost or con-
densation on the inner liner when insulation is placed only
between the sub-girt lines. The metal-to-metal contact (out-
side sheet-sub-girt-inside sheet) should be broken to reduce
thermal bridging. A detail that has been used successfully
is shown in Figure 5-1. Another option is to provide rigid
insulation between the girt and liner on one side. In any
event, the wall should be evaluated for thermal transmittance
in accordance with Energy Efficient Design of New Build-
ings Except Low-Rise Residential Buildings, ASHRAE 90.1
(ASHRAE, 2013).
5.2 FACTORY-ASSEMBLED PANELS
Factory-assembled panels typically consist of interior liner
panels, exterior metal panels, and insulation. Panels provid-
ing various insulating values are available from several man-
ufacturers. These systems are generally proprietary and must
be designed according to manufacturer’s recommendations.
The particular advantages of these factory-assembled pan-
els are:
1. Panels are lightweight and do not require heavy erec-
tion cranes, large foundations, or heavy spandrels.
2. Panels can have a hard surface interior liner.
3. Panel sidelap fasteners are typically concealed, pro-
ducing a “clean” appearance.
4. Documented panel performance characteristics deter-
mined by test or experience may be available from
manufacturers.

Disadvantages of factory-assembled panels include:
1. Once a choice of panel has been made, future expan-
sions may effectively require use of the same panel
to match color and profile, thus competition is essen-
tially eliminated.
2. Erection procedures usually require starting in one
corner of a structure and proceeding to the next cor-
ner. Due to the interlocking nature of the panels it may
be difficult to add openings in the wall.
3. Close attention to coordination of details and toler-
ances with collateral materials is required.
4. Thermal changes in panel shape may be more
apparent.
5.3 PRECAST WALL PANELS
Precast wall panels for industrial buildings could utilize one
or more of a variety of panel types including:
1. Hollow core slabs
2. Double-tee sections
3. Site cast tilt-up panels
4. Factory-cast panels
Panels can be either load bearing or nonload bearing and
can be obtained in a wide variety of finishes, textures, and
colors. Also, panels may be of sandwich construction and
contain rigid insulation between two layers of concrete.
Such insulated panels can be composite or noncomposite.
Composite panels typically have a positive concrete connec-
tion between inner and outer concrete layers. These panels
are structurally stiff and are good from an erection point of
view, but the positive connection between inner and outer
layers may lead to exterior surface cracking when the panels
are subjected to a temperature differential. The direct con-
nection can also provide a path for thermal bridging, which
may be a problem in high-humidity situations.
True sandwich panels connect inner and outer concrete
layers with flexible metal ties. Insulation is exposed at all
panel edges. These panels are more difficult to handle and
erect but typically perform well.
Erecting precast wall panels may be problematic. Lifting
lugs cast into the top of the panels are intended for vertical
lifting. When lifting from a horizontal shipping or storage
condition, the area around the lug may rupture, causing a
safety hazard and damaging the panel.
Precast panels have multiple advantages for use in indus-
trial buildings:
1. A hard surface is provided inside and out.
Fig. 5-1. Wall thermal break detail.

2. These panels produce an architecturally “clean”
appearance.
3. Panels have inherent fire-resistance characteristics.
4. Intermediate girts are typically not required.
5. Use of load-bearing panels can eliminate exterior
framing and reduce cost.
6. They provide an excellent sound barrier.
Disadvantages of precast wall panel systems include:
1. Matching colors of panels in future expansion may be
difficult.
2. Composite sandwich panels have “cold spots” with
potential condensation problems at panel edges.
3. Adding wall openings can be difficult.
4. Panels have poor sound-absorption characteristics.
5. Foundations and grade beams may be heavier than for
other panel systems.
6. Heavier eave struts are required for steel-framed
structures than for other systems.
7. Heavy cranes are required for panel erection.
8. If panels are used as load-bearing elements, expan-
sion in the future could present problems.
9. Close attention to tolerances and details to coordinate
divergent trades are required.
10. Added dead weight of walls can affect seismic design.
5.4 MASONRY WALLS
Use of masonry walls in industrial buildings is common.
Walls can be load bearing or nonload bearing.
Some advantages of the use of masonry wall construction
are:
1. A hard surface is provided inside and out.
2. Masonry walls have inherent fire resistance
characteristics.
3. Intermediate girts are typically not required.
4. Use of load-bearing walls can eliminate exterior fram-
ing and reduce cost.
5. Masonry walls can serve as shear walls to brace col-
umns and resist lateral loads.
6. Walls produce a flat finish, resulting in ease of both
maintenance and dust-control considerations.
Disadvantages of masonry include:
1. Masonry has comparatively low material bend-
ing resistance. Walls are typically adequate to resist
normal wind loads, but interior impact loads can
cause damage.
2. Foundations may be heavier than for metal wall-panel
construction.
3. Special consideration is required in the use of masonry
ties, depending on whether the masonry is erected
before or after the steel frame.
4. Buildings in seismic regions may require special rein-
forcing and added dead weight may increase seismic
forces.
5.5 GIRTS
Typical girts for industrial buildings are hot-rolled chan-
nel sections or cold-formed steel C- or Z-shaped sections.
In some instances, HSS are used to eliminate the need for
compression-flange bracing. In recent years, cold-formed
sections have gained popularity because of their low cost.
As mentioned earlier, cold-formed Z-shaped sections can
be easily lapped to achieve continuity, resulting in further
weight savings and reduced deflections. Z-shaped sections
also ship economically. Additional advantages of cold-
formed sections compared with hot-rolled girt shapes are:
1. Metal wall panels can be attached to cold-formed
girts quickly and inexpensively using self-drilling
fasteners.
2. The use of sag rods is often not required.
Hot-rolled girts are often used when:
1. Corrosive environments dictate the use of thicker
sections.
2. Common cold-formed sections do not have sufficient
strength for a given span or load condition.
3. Girts will receive substantial abuse from operations.
4. Designers are unfamiliar with the availability and
properties of cold-formed sections.
5. In some instances, the overall cost of the erected girt
system using hot-rolled sections can be competitive
with cold-formed girts depending on the fabricators
equipment and the ability of the erector to panelize
the wall system for erection.
Both hot-rolled and cold-formed girts subjected to wind-
pressure loads are typically considered laterally braced by
the wall sheathing. Negative moment regions in continuous
cold-formed girt systems are typically considered laterally
braced at inflection points and at girt-to-column connec-
tions. Continuous systems can be analyzed by assuming
either a single prismatic section throughout or a double
moment of inertia condition within the lapped section of

the cold-formed girt. Research indicates that an analytical
model assuming a single prismatic section is closer to exper-
imentally determined behavior (Robertson and Kurt, 1986).
The use of sag rods is typically required to maintain hori-
zontal alignment of hot-rolled sections. The sag rods are
often used to provide lateral restraint against buckling for
internal suction loads. When used as bracing, the sag rods
must be designed to take tension in either the upward or
downward direction. The wall paneling is assumed to pro-
vide lateral support for external pressure loads. Lateral sta-
bility for the girt based on this assumption is checked using
AISC Specification Chapter F.
The typical design procedure for hot-rolled girts is as
follows:
1. Select the girt size based on external pressure loads,
assuming full-flange lateral support.
2. Check the selected girt for sag rod requirements based
on deflections and bending stresses about the weak
axis of the girt.
3. Check the girt for internal suction loads using AISC
Specification Chapter F.
4. If the girt is inadequate, increase its size or add sag
rods.
5. Check the girt for serviceability requirements.
6. Check the sag rods for their ability to resist the twist
of the girt due to suction loads. The sag rod and siding
act to provide the torsional brace.
Cold-formed girts should be designed in accordance with
the provisions of the AISI Specification. Many manufactur-
ers of cold-formed girts will provide design assistance and
offer load span tables to aid design.
AISI Specification Section I6.2.3, “Compression Mem-
bers Having One Flange Through-Fastened to Sheath-
ing,” provides a means for determining cold-formed girt
strength when the compression flange of the girt is attached
to sheeting (fully braced). For lapped systems, the sum of
the moment capacities of the two lapped girts is normally
assumed to resist the negative moment over the support. For
full continuity to exist, a lap length on each side of the col-
umn support should be equal to at least 1.5 times the girt
depth (Robertson and Kurt, 1986). Additional provisions are
given in AISI Specification Section G for strength consider-
ations relative to shear, web crippling, and combined bend-
ing and shear.
AISI Specification Section I6.2.1, “Flexural Members
Having One Flange Through-Fastened to Deck or Sheath-
ing,” provides a simple procedure to design cold-formed
girts subjected to suction loading. The basic equation for the
determination of the girt strength is:
M RS F
n e y
= (5-1)
The values of R are shown in Table 5-1:
Table 5-1. Simple Span
C- or Z-Shaped Section R Values
Depth Range,
in. Profile R
d ≤ 6.5 C or Z 0.70
6.5 d ≤ 8.5 C or Z 0.65
8.5 d ≤ 12 Z 0.50
8.5 d ≤ 12 C 0.40
where
Fy = specified minimum yield stress, ksi
Se =
elastic section modulus, of the effective section,
calculated at the extreme compression or tension
fiber at Fy, in.3
Other restrictions relative to insulation, girt geometry,
wall panels, and fastening systems between wall panels and
girts are discussed in the AISI Specification. It should also
be mentioned that consideration should be given to tolerance
differences between erected columns and girts. The use of
slotted holes in girt-to-column attachments is often required.
5.6 WIND COLUMNS
When bay spacing exceeds 30 ft, additional intermediate
columns may be required to provide for economical girt
design. Two considerations that should be emphasized are:
1. Sufficient bracing of the wind columns to accommo-
date wind suction loads is needed. This is typically
accomplished by bracing the interior flanges of the
columns with angles that connect to the girts.
2. Proper attention should be paid to the connection
at the top of the columns. For intermediate sidewall
columns, secondary roof-framing members must
be provided to transfer the wind reaction at the top
of the column into the roof-bracing system. Do not
rely on “trickle theory” (i.e., a force will find a way
to trickle out of the structure.). A positive and calcu-
lable system is necessary to provide a traceable load
path, such as the diagonals and angle struts shown in
Figure 5-2. Bridging systems or bottom-chord exten-
sions on joists can be used to dissipate these forces,
but the stresses in the system must be checked. If the
wind columns have not been designed for axial load,
a slip connection would be necessary at the top of the
column.
Small wind reactions can be transferred from the wind
columns into the roof diaphragm system as shown in
Figure 5-2.

Allowable values for attaching metal deck to structural
members can be obtained from screw manufacturers. Allow-
able stresses in welds to metal deck can be determined from
Structural Welding Code—Steel, AWS D1.1/D1.1M:2015
(AWS, 2015), or from the AISI Specification. In addition to
determining the fastener requirements to transfer the con-
centrated load into the diaphragm, the designer must also
check the roof diaphragm for its strength and stiffness. This
can be accomplished by using the procedures contained in
the SDI Diaphragm Design Manual.
Fig. 5-2. Wind column reaction load transfer.

Chapter 6
Framing Schemes
The selection of the best framing scheme for an industrial
building without cranes is dependent on numerous consid-
erations and often depends on the owner’s requirements. It
may not be possible to give a list of rules by which the best
scheme can be assured. If “best” means low initial cost, then
the owner may face major expenses in the future for opera-
tional expenses or problems with expansion. Extra dollars
invested at the outset reduce potential future costs.
The economics of use of long-span versus short-span
joists and purlins has been mentioned previously in this
Guide. This section expands on the selection of the main
framing system. No attempt has been made to evaluate foun-
dation costs. In general, if a deep foundation system (e.g.,
piles or drilled piers) is required, longer bay spacing is typi-
cally more economical.
The consideration of bay sizes must include not only roof
and frame factors but also the wall system. The cost of large
girts and thick wall panels may cancel the savings antici-
pated if the roof system alone is considered.
AISC offers a publication that may aid in the design of
efficient framing entitled Detailing for Steel Construction
(AISC, 2009).
6.1 BRACED FRAMES VERSUS RIGID FRAMES
The design of rigid frames is explained in numerous text-
books and professional journals and will not be covered
here; however, a few concepts will be presented concerning
the selection of a braced-frame versus a rigid-frame struc-
tural system. There are several situations for which a rigid
frame system is likely to be superior.
1. Braced frames may require bracing in both walls and
roof. Bracing frequently interferes with plant opera-
tions and future expansion. If either consideration is
important, a rigid frame structure may be the answer.
2. The bracing of a roof system can be accomplished
through X-bracing or a roof diaphragm. In either
case, the roof becomes a large horizontal beam span-
ning between the walls or bracing that must transmit
the lateral loads to the foundations. For large span-
to-width ratios (greater than 3:1), the bracing require-
ments become excessive. A building with dimensions
of 100 ft by 300 ft with potential future expansion in
the long direction may best be suited for rigid frames
to minimize or eliminate bracing that would interfere
with future changes.
Experience has shown that there are situations when
braced-frame construction may prove to be more economi-
cal than either standard metal building systems or special
rigid-frame construction if certain sacrifices on flexibility
are acceptable.
Use of a metal building system requires a strong interac-
tion between the designer and the metal building manufac-
turer. Much of the detailing process concerning the design
is provided by the manufacturer, and the options open to the
buyer may reflect the limits of the manufacturer’s standard
product line and details.
6.2 HSS VERSUS W-SHAPE COLUMNS
The design of columns in industrial buildings includes con-
siderations that do not apply to other types of structures.
Interior columns can normally be braced only at the top and
bottom; thus, square HSS columns are desirable due to their
equal stiffness about both principal axes. Difficult connec-
tions with HSS members can be eliminated in single-story
frames by placing the beams over the tops of the HSS. Thus,
a simple-to-fabricate cap plate detail with bearing stiffen-
ers on the girder web may be designed. Other advantages of
HSS columns include the fact that they require less paint than
equivalent W-shapes, and they are pleasing aesthetically.
W-shapes may be more economical than HSS for exterior
columns for the following reasons:
1. The wall system (girts) may be used to brace the weak
axis of the column. It should be noted that a stiffener
or brace may be required for the column if the inside
column flange is in compression and the girt connec-
tion is assumed to provide a braced point in design.
2. Bending moments due to wind loads are predominant
about one axis.
3. It is easier to frame girt connections to a W-shape
than to an HSS section. Because HSS have no flanges,
extra clip angles are required to connect girts.
6.3 MEZZANINE AND PLATFORM FRAMING
Mezzanines and platforms are often required in industrial
buildings. Design considerations are dictated by the type of
usage. For proper design the designer needs to consider the
following design parameters:
1. Occupancy or use
2. Design loads (uniform and concentrated)
3. Deflection criteria

4. Surface type
a. Raised pattern plate
b. Smooth plate
c. Concrete composite slab
d. Concrete noncomposite slab
e. Hollow core slabs (topped or untopped)
f. Plywood
g. Guard rail requirements, including removable
sections
h. Future expansion
i. Vibration control
j. Lateral stability requirements
6.4 ECONOMIC CONSIDERATIONS
As previously mentioned, bay sizes and column spacing are
often dictated by the function of the building. Economics,
however, should also be considered. In general, as bay sizes
increase, the weight of the horizontal framing increases.
This will mean additional cost unless offset by savings in
foundations or erection. Studies have indicated that square
or slightly rectangular bays usually result in more economi-
cal structures.
In order to evaluate various framing schemes, a prototype
general merchandise structure was analyzed using various
spans and component structural elements. The structure was
a 240-ft × 240-ft building with a 25-ft eave height. The total
roof load was 48 psf, and beams with Fy = 50 ksi were used.
Columns were HSS with a yield strength of 50 ksi. Only
gravity loads were considered.
Variables in the analysis were:
1. Joist spans: 25 ft, 30 ft, 40 ft, 50 ft, and 60 ft
2. Girder spans, W-shapes: 25 ft, 30 ft, 40 ft, 48 ft, and
60 ft
Table 6-1. Relative Costs*
Joist Data Main Framing W-Shape Members
Depth, in. Span, ft
Span, ft
25 30 40 48 60
16 25 1.10 1.10 1.25 1.31 1.53
18 30 1.12 1.07 1.20 1.28 1.50
24 40 1.16 1.05 1.15 1.28 1.47
30 50 1.22 1.18 1.20 1.30 1.54
32 60 1.33 1.30 1.30 1.33 1.60
*
Cost included fabrication and erection of primary and secondary framing (no deck). A total gravity load of 48 psf was
used in all designs. Uplift and lateral bracing requirements were not included.
Cost data were determined from several fabricators.
The data did not include sales tax or shipping costs. The
study yielded several interesting conclusions for engineers
involved in industrial building design. An examination of the
tabular data in Table 6-1 shows that the most economical
framing scheme was the one with beams spanning 30 ft and
joists spanning 40 ft.
Another factor that may be important is that for the larger
bays (greater than 30 ft), normal girt construction becomes
less efficient using C- or Z-shaped sections without inter-
mediate wind columns being added. For the 240-ft × 240-ft
building being considered, wind columns could add $0.10
per square foot of roof to the cost. Interestingly, if the
building were 120 ft × 120 ft, the addition of intermediate
wind columns would add $0.20 per square foot because the
smaller building has twice the perimeter-to-area ratio com-
pared to the larger structure.
Additional economic and design considerations are as
follows:
1. When steel joists are used in the roof framing it is
generally more economical to span the joists in the
long direction of the bay.
2. K-series joists are more economical than LH joists;
thus, an attempt should be made to limit spans to
those suitable for K-series joists.
3. For 30-ft to 40-ft bays, efficient framing may consist
of continuous or double-cantilevered girders sup-
ported by columns in one direction and joists span-
ning the other direction.
4. If the girders are continuous, plastic design is often
used. Connection costs for continuous members may
be higher than for cantilever design; however, a plasti-
cally designed continuous system will have superior
behavior when subjected to pattern load cases. All
flat-roof systems must be checked to prevent ponding
problems. See Section 3.5.

5. Simple-span rolled beams are often substituted for
continuous or double-cantilevered girders where
spans are short. The simple-span beams often have
adequate flexural strength. The connections are sim-
ple, and the savings from easier erection of such sys-
tems may overcome the cost of any additional weight.
6. For large bay dimensions in both directions, a popu-
lar system consists of cold-formed or hot-rolled steel
purlins or joists spanning 20 ft to 30 ft to secondary
trusses spanning to the primary trusses. This framing
system is particularly useful when heavily loaded
monorails must be hung from the structure. The sec-
ondary trusses in conjunction with the main trusses
provide excellent support for the monorails.
7. Consideration must be given to future expansion and/
or modification, where columns are either moved or
eliminated. Such changes can generally be accom-
plished with greater ease where simple-span condi-
tions exist.

Chapter 7
Bracing Systems
7.1 RIGID-FRAME SYSTEMS
There are many considerations involved in providing lateral
stability to industrial structures. If a rigid frame is used, lat-
eral stability parallel to the frame is provided by the frame.
However, for loads perpendicular to the main frames and for
wall bearing and “post and beam” construction, lateral brac-
ing is not inherent and must be provided. It is important to
reemphasize that future expansion may dictate the use of a
rigid frame or a flexible (movable) bracing scheme.
Because industrial structures are typically light and low
in profile, wind and seismic forces may be relatively low.
Rigid-frame action can be easily and safely achieved by pro-
viding a properly designed member at a column line. If joists
are used as part of the rigid frame, the designer is cautioned
on the following points:
1. The design loads must be given on the structural
plans so that the proper design can be provided by
the joist manufacturer. The procedure must be used
with conscious engineering judgment and full rec-
ognition that standard steel joists are designed as
simple-span members subject to distributed loads.
(See the SJI Specification). Bottom chords are typi-
cally sized for tension only. The simple attachment of
the bottom chord to a column to provide lateral stabil-
ity will cause gravity load end moments that cannot be
ignored. The designer should not try to select member
sizes for these bottom chords because each manufac-
turer’s design is unique and proprietary.
2. It is necessary for the designer to provide a well-
designed connection to both the top and bottom
chords to develop the induced moments without caus-
ing excessive secondary bending moments in the joist
chords.
3. The system must have adequate stiffness to prevent
drift related problems such as cracked walls and parti-
tions, broken glass, leaking walls and roofs, and mal-
functioning or inoperable overhead doors.
7.2 BRACED SYSTEMS
7.2.1 Roof Diaphragms
The most economical roof bracing system is achieved by use
of a steel deck diaphragm. The deck is provided as the roof-
ing element, and the effective diaphragm is obtained at little
additional cost for extra deck connections. A roof diaphragm
used in conjunction with wall X-bracing or a wall diaphragm
system is probably the most economical bracing system that
can be achieved. Diaphragms are most efficient in relatively
square buildings; however, an aspect ratio up to three can be
accommodated. A cold-formed steel diaphragm is analogous
to the web of a plate girder. That is, its main function is to
resist shear forces. The perimeter members of the diaphragm
serve as the “flanges.”
The design procedure is quite simple. The basic param-
eters that control the strength and stiffness of the diaphragm
are:
1. Profile shape
2. Deck material thickness
3. Span length
4. The type and spacing of the fastening of the deck to
the structural members
5. The type and spacing of the sidelap connectors
The profile, thickness, and span of the deck are typically
based on gravity load requirements. The type of fastening
(i.e., welding, screws, or power-driven pins) is often based
on the designer’s or contractor’s preference. Thus, the main
design variable is the spacing of the fasteners. The designer
calculates the maximum shear per foot of diaphragm and
then selects the fastener spacing from the load tables. Load
tables are most often based on the requirements set forth
in the SDI Diaphragm Design Manual and the AISI North
American Standard for the Design of Profiled Steel Dia-
phragm Panels (AISI, 2016b), hereafter referred to as the
AISI Design of Profiled Steel Diaphragm Panels.
Deflections are calculated and compared with serviceabil-
ity requirements. The calculation of flexural deformations is
handled in a conventional manner. Shear deformations can
be obtained mathematically, using shear deflection equations
if the shear modulus of the formed deck material making up
the diaphragm is known. Deflections can also be obtained
using empirical equations such as those found in the SDI
Diaphragm Design Manual and the AISI Design of Profiled
Steel Diaphragm Panels. In addition, most metal deck manu-
facturers publish tables in which strength and stiffness (or
flexibility) information is presented. In order to illustrate the
diaphragm design procedure, a design example is presented
in the following text. The calculations presented are based
on SDI’s procedure.

7.2.1.1—Diaphragm Design Example
Given:
Design the roof diaphragm for the plan shown in Figure 7-1. Use 20-gage (0.0358-in.), 12-in.-deep intermediate rib deck span-
ning 5 ft 6 in. to support the gravity loads. Steel joists span in the north-south direction. Use welds to connect the deck to the
structural members and #10 screws for the sidelaps.
The nominal eave wind loads for LRFD and ASD are shown in Figure 7-1. Note that the length-to-width ratio of the diaphragm
does not exceed 3, which is the typically accepted maximum for diaphragms.
Solution:
1. Calculate the maximum diaphragm shear.
LRFD ASD
Shear:
V
WL
2
323 lb/ft 208 ft
2
33,600 lb
r
( )( )
=
=
=
Shear per ft:
v
V
span
33,600 lb
96 ft
350 lb/ft
r
r
=
=
=
Shear:
V
WL
2
250 lb/ft 208 ft
2
26,000 lb
r
( )( )
=
=
=
Shear per ft:
v
V
span
26,000 lb
96 ft
271 lb/ft
r
r
=
=
=
2. Obtain the shear capacity of the deck from the SDI Diaphragm Design Manual.
For a 20-gage (0.0358-in.), 12-in.-deep intermediate rib deck, spanning 5 ft 6 in. with #10 sidelap screws, ϕ = 0.70, Ω = 2.35.
The shear values from the SDI Diaphragm Design Manual are given in Table 7-1.
Use connector patterns as shown in Figure 7-2 for ASD.
3. Calculate the lateral deflection of the diaphragm (using 36/4 weld pattern and one sidelap screw). The deflection equations
from the SDI Diaphragm Design Manual, Section 10, are:
a. For bending:
wL
EI
5
384
b
4
=
Δ

(7-1)
b. For shear:
wL
DG
8
s
2
=
Δ
′
(7-2)
where
D = diaphragm depth, ft
G′ = shear stiffness as determined from SDI Diaphragm Design Manual tables, lb/in.
L = diaphragm span, ft
w = eave force, kip/ft

Fig. 7-1. Diaphragm plan.
Fig. 7-2. Roof load diagram (ASD shears shown).
Table 7-1. Diaphragm Shear Values
Connection Pattern
Nominal Shear
Strength
Design Shear Strength
LRFD ASD
Vn, lb/ft ϕVn, lb/ft Vn/Ω, lb/ft
36/4 weld pattern and one sidelap screw 615 431 262
36/4 weld pattern and two sidelap screws 715 501 304
36/5 weld pattern and one sidelap screw 785 550 334

From the SDI Diaphragm Design Manual, Section 9:
Dxx = Dir (intermediate rib)
= 645 ft
K1 = 0.561 ft-1
K2 = 1,056 kip/in.
K4 = 3.45
For a bare deck:
G
K
K
D
L
K L
0.3
3
1,056 kip/in.
3.45
0.3 645 ft
5.5 ft
3 0.561 ft 5.5 ft
22.1 kip/in.
xx
v
v
2
4 1
1
( )
( )
( )
( )
( )
′ =
+
⎛
⎝
⎜
⎞
⎠
⎟ +
=
+
⎡
⎣
⎢
⎤
⎦
⎥ +
=
−

(7-3)
The moment of inertia, I, can be based on an assumed area of the perimeter member. Assuming the edge member has an area of
3.00 in.2
, the moment of inertia is:
I Ad
2
2 3.00 in. 48 ft 12 in./ft
1.99 10 in.
2
2 2
6 4
( )[ ]
( )( )
=
=
= ×
(7-4)
where
d = distance from centroid to outer edge, in.
The bending deflection can then be calculated as:
( )
( )( ) ( )
( )
=
Δ
=
×
=
wL
EI
5
384
5 0.250 kip/ft 208 ft 12 in./ft
384 29,000 ksi 1.99 10 in.
0.182 in.
b
4
4 3
6 4

(7-1)
The shear deflection can then be calculated as:
wL
DG
8
0.250 kip/ft 208 ft
8 96 ft 22.1 kip/in.
0.637 in.
s
2
2
( )
( )
( )
( )
=
Δ
′
=
=
(7-2)
And the total deflection is:
0.182 in. 0.637 in.
0.819 in.
b s
= +
Δ Δ
Δ
= +
=
(7-5)

To transfer the shear forces into the east and west walls of the structure, the deck can be welded directly to the perimeter beams.
The deck must be connected to the perimeter beams with the same number of fasteners as required in the field of the diaphragm.
The reader is cautioned regarding connecting steel deck to the end walls of buildings. If the deck is to be connected to a shear
wall and a joist is placed next to the wall, allowance must be made for the camber in the edge joist in order to connect the deck
to the wall system. If proper details are not provided, diaphragm connection may not be possible, and field adjustments may be
required. Where the edge joist is eliminated near the endwall, the deck can often be pushed down flat on an endwall support. If
the joist has significant camber, it may be necessary to provide simple-span pieces of deck between the wall and the first joist.
A heavier deck thickness may be required due to the loss in continuity. The edge should be covered with a sheet metal cap to
protect the roofing materials. This can present an additional problem because the sharp edge of the deck will stick up and pos-
sibly damage the roofing.
Along the north and south walls, a diaphragm chord can be provided by attaching an angle to the top of the joists as shown in
Figure 7-3. The angle also stiffens the deck edge and prevents tearing of roofing materials along the edge where no parapet is
provided under foot traffic. In some designs an edge angle may also be required for the sidelap connections for wind forces in
the east-west direction. Also, shear connectors may be required to transfer these forces into the perimeter beam. A typical shear
collector is shown in Figure 7-4.
Fig. 7-3. Rake angle.
Fig. 7-4. Tube or filler channel.

temporary bracing and in establishing the need for such
bracing.
Secondly, the AISC Code of Standard Practice does not
emphasize that the process of erection can induce forces
and stresses into components and systems such as footings
and piers that are not part of the structural steel framework.
Unless otherwise specified in the contract documents, it is
the practice of architects and engineers to design the ele-
ments and systems in a building for the forces acting upon
the completed structure only. An exception to this is the
requirement in OSHA Subpart R that column bases be
designed to resist a 300-lb downward force acting at 18 in.
from the face of columns.
Without a detailed erection bracing plan, it is difficult
for anyone in the design/construction process to evaluate
the performance of the erector relative to bracing without
becoming involved in the process itself. This is inconsis-
tent with the determination that temporary bracing is the
sole responsibility of the erector. The lack of emphasis on
the necessity that the erector check the effect of erection-
induced forces on other elements has, at times, allowed erec-
tion problems to be erroneously interpreted as having been
caused by other reasons. This is most obvious in the erection
of steel columns.
To begin and continue the erection of steel framework, it
is necessary to erect columns first. This means that at one
time or another, each building column is set in place without
stabilizing framing attached to it in two perpendicular direc-
tions. Without such framing, the columns must cantilever for
a time from the supporting footing or pier unless they are
braced by adequate guys or unless the columns and beams
are designed and constructed as rigid frames in both direc-
tions. The forces induced by the cantilevered column on the
pier or footing may not have been considered by the build-
ing designer unless this had been specifically requested. It
is incumbent upon the steel erector to make a determination
of the adequacy of the foundation to support cantilevered
columns during erection.
Trial calculations suggest that large forces can be induced
into anchor rods, piers, and footings by relatively small
forces acting at or near the tops of columns. Also, wind
forces can easily be significant, as can be seen in the fol-
lowing example. Figure 7-5 shows a section of an unbraced
frame consisting of three columns and two beams. The
beams ends are pinned. Wind forces are acting perpendicular
to the frame line.
Using a shape factor of 2.0 for a 40-mph wind directed
at the webs of the W12 columns, a service-level base
moment of approximately 18 kip-ft occurs. If a 5-in. × 5-in.
placement pattern were used with four anchor rods and an
ungrouted base plate, a tensile force of approximately 21.6
kips would be applied to the two anchor rods. The allowable
tensile strength for a w-in. ASTM F1554 (ASTM, 2019)
Grade A36 anchor rod is 9.6 kips. Even if the bolts were
7.2.2 Roof X-Bracing
An alternative to using the roof deck as the roof diaphragm
is to use cross bracing to develop a horizontal truss system.
As with the metal deck diaphragm, as the length-to-width
ratio of the building becomes larger than 3-to-1, the diagonal
forces in the truss members may require consideration of an
alternate bracing method.
An especially effective way to develop a cross-braced roof
is to utilize flat bar stock resting on the roof joists. The use
of 4-in.-thick bar stock does not usually interfere with deck
placement and facilitates erection.
7.2.3 Vertical Bracing
In braced buildings, the roof diaphragm loads or the roof
cross-bracing loads are transferred to a vertical braced frame,
which in turn transfers the loads to the foundation level. In
most cases, the vertical bracing is located at the perimeter
of the structure so as not to interfere with plant operations.
The vertical bracing configuration most frequently used is
a cross-braced system using angles or rods designed only
to function as tension members. However, in areas of high
seismicity, a vertical bracing system that incorporates
tension/compression members is often required. In these
cases, other bracing forms may be used, such as chevron
bracing or eccentrically braced frames. In buildings with
large aspect ratios, bracing may be required in internal bays
to reduce the brace forces and to reduce foundation overturn-
ing forces.
7.3 TEMPORARY BRACING
Proper temporary bracing is essential for the timely and safe
erection and support of the structural framework until the
permanent bracing system is in place. The need for tempo-
rary bracing is recognized in AISC Specification Section
M4.2 and in AISC Code of Standard Practice Section 7.10.
The AISC Code of Standard Practice places the responsi-
bility for temporary bracing solely with the erector. This is
appropriate because temporary bracing is an essential part
of the work of erecting the steel framework. While the gen-
eral requirements of the AISC Code of Standard Practice are
appropriate in establishing the responsibility for temporary
erection bracing, two major issues have the potential to be
overlooked in the process.
First, it is difficult to judge the adequacy of temporary
bracing in any particular situation using only the general
requirements as a guide. There is no codified standard that
can be applied in judging whether or not a minimum level of
conformity has been met. However, Design Loads on Struc-
tures During Construction, ASCE 37-14 (ASCE, 2014),
and AISC Design Guide 10, Erection Bracing of Low-Rise
Structural Steel Buildings (Fisher and West, 1997), can be
useful in making evaluations of the adequacy of proposed

stability should have the necessary elements providing the
stability identified in the contract documents along with the
schedule of their completion. The coordination of the instal-
lation of such elements is a matter that must be addressed by
the general contractor.
In some instances, the steel frame is providing lateral
support for other elements such as masonry walls. In these
cases, it may be necessary to install and leave temporary
bracing for the masonry in place until the mortar has cured
and the permanent support for the wall is erected.
Temporary support beyond the requirements discussed in
the preceding text would be the responsibility of the owner
according to theAISC Code of Standard Practice. For exam-
ple, if the steel frame and its temporary bracing are to support
other nonstructural elements, the responsibility for this must
be clearly identified and the reactions from the elements are
to be provided to the erector. Otherwise the responsibility for
this falls to others, not the erector.
The timing of column base grouting affects the perfor-
mance of column bases during erection. The AISC Code of
Standard Practice establishes the timing of grouting and
assigns the responsibility for grouting to the owner. The
erector should be aware of the schedule for this work.
All of the foregoing draws attention to the need for care,
attention, and thoroughness on the part of the erector in
preparing and following a temporary bracing and erection
scheme.
fully in the concrete, they would be severely overstressed
and would likely fail. Four 18-in. anchor rods would be
required to resist the wind force.
Of course, not only the size of the anchor rod is affected,
but the design of the base plate and its attachment to the
column, the spacing of the anchor rods, and the design of the
pier and footing must also be checked.
Guying can also induce forces into the structure in the
form of base shears and uplift forces. These forces may not
have been considered in the design of the affected members.
This must also be checked by the erector. The placement of
material such as decking on the incomplete structure can
induce unanticipated loadings. This loading must also be
considered explicitly. OSHA Subpart R states that no deck-
ing bundles may be placed on the frame until a qualified
person has documented that a structure or portion of it is
capable of supporting the load.
Erection bracing involves other issues as well. First, the
AISC Code of Standard Practice distinguishes between
frames in which the frame is stabilized by construction in
the control of the erector versus those frames in which other
nonstructural steel elements are required for the stability of
the frame. The distinction is drawn because the timing of the
removal of bracing is affected. In a structural steel frame,
where lateral stability is achieved in the design and detailing
of the framework itself, the bracing can be removed when
the erector’s work is complete. Steel framework that relies
on elements other than the structural steel to provide lateral
Fig. 7-5. Erection bracing example.

Chapter 8
Column Anchorage
Building columns must be anchored to the foundation sys-
tem to transfer tensile forces, shear forces, and overturning
moments. This Design Guide is limited to the design of
column anchorage for tensile forces. The principles dis-
cussed here can be applied to the design of anchorage for
overturning moments. The reader is referred to Base Plate
and Anchor Rod Design, AISC Design Guide 1 (Fisher and
Kloiber, 2006), hereafter referred to as AISC Design Guide
1, for an extensive treatment of base plates and anchor rod
design plus methods of resisting shear loads.
Improper design, detailing, and installation of anchor rods
has caused numerous structural problems in industrial build-
ings. These problems include:
1. Inadequate sizing of the anchor rods.
2. Inadequate development of the anchor rods for tension.
3. Inadequate design or detailing of the foundation for
forces from the anchor rods.
4. Inadequate base plate thickness.
5. Inadequate design and/or detailing of the anchor rod-
to-base plate interface.
6. Misalignment or misplacement of the anchor rods
during installation.
7. Fatigue.
The reader should be familiar with the OSHA require-
ments contained in Safety and Health Standards for the
Construction Industry, 29 CFR Part 1926 Part R Safety
Standards for Steel Structures (OSHA, 2010b). This docu-
ment was partially produced to prevent construction acci-
dents associated with column base plates. For example, it
requires that all column bases have four anchor rods. The
following discussion presents methods of designing and
detailing column bases.
8.1 RESISTING TENSILE FORCES WITH
ANCHOR RODS
The design of anchor rods for tension consists of four steps:
1. Determine the maximum net uplift for the column.
2. Select the anchor rod material and number and size of
anchor rods to accommodate this uplift.
3. Determine the appropriate base plate size, thickness,
and welds to transfer the uplift forces. Refer to AISC
Design Guide 1.
4. Determine the method for developing the anchor rod
in the concrete (i.e., transferring the tensile force from
the anchor rod to the concrete foundation).
Step 1
The maximum net uplift for the column is obtained from the
structural analysis of the building for the prescribed building
loads. The use of light metal roofs on industrial buildings is
very popular. As a result of this, the uplift due to wind often
exceeds the dead load; thus, the supporting columns are sub-
jected to net uplift forces. In addition, columns in rigid bents
or braced bays may be subjected to net uplift forces due to
overturning.
Step 2
Anchor rods should be specified to conform to ASTM
F1554. Grades 36, 55, and 105 are available in this specifi-
cation where the grade number represents the yield stress of
the anchor. Unless otherwise specified, the end of the anchor
will be color coded to identify its grade. Welding is permit-
ted to Grade 36 and Grade 55 if it conforms to the ASTM
F1554 Supplementary Requirement S1.
Anchor rods should no longer be specified to ASTM A307
(ASTM, 2019) even if the intent is to use the A307 Grade
C anchor rod that conforms to ASTM A36 (ASTM, 2019)
properties. Anchor rods conforming to the ASTM specifica-
tions listed for anchor rods and threaded rods in AISC Speci-
fication Section A3.4 can be used.
The number of anchor rods required is a function of the
maximum net uplift on the column and the allowable ten-
sile load per rod for the anchor rod material chosen. Prying
forces in anchor rods are typically neglected. This is usually
justified when the base plate thickness is calculated assum-
ing cantilever bending about the web and/or flange of the
column section (as described in step 3). However, calcula-
tions have shown that prying forces may not be negligible
when the rods are positioned outside the column profile and
the rod forces are large. A conservative estimate for these
prying forces can be obtained using a method similar to that
described for hanger connections in the AISC Steel Con-
struction Manual (AISC, 2017), hereafter referred to as the
AISC Manual.
Another consideration in selection and sizing of anchor
rods is fatigue. For most building applications where uplift
loads are generated from wind and seismic forces, fatigue

Step 4
For the transfer of forces to the concrete foundation, AISC
Specification Section J9 defers to Building Code Require-
ments for Structural Concrete and Commentary, ACI 318-14
(ACI, 2014). ACI 318-14, Chapter 17, addresses the anchor-
ing to concrete of cast-in or post-installed expansion or
undercut anchors. In the method, the concrete cone is con-
sidered to be formed at an angle of approximately 35° (1 to
1.5 slopes). For simplification of application, the cone is
considered to be square rather than round in plan.
The concrete breakout stress, ft, in the method is consid-
ered to decrease with increase in size of the breakout surface.
Consequently, the increase in strength of the breakout in the
method is proportional to the embedment depth to the power
of 1.5 (or to the power of 5
3 for deeper embedment).
In many cases, it is necessary to use reinforcement to
anchor the breakout cone in order to achieve the shear capac-
ity as well as the ductility desired.
Careful consideration should be given to the size of the
anchor rod holes in the base plate when transferring shear
forces from the column base plate to the anchor rods. The
designer should use the recommended anchor rod hole
diameters and minimum washer diameters as shown in
AISC Manual Table 14-2. These recommended hole sizes
vary with rod diameter and are considerably larger than nor-
mal bolt hole sizes. If slip of the column base before bear-
ing against the anchor rods is of concern, then the designer
should consider using plate washers between the base plate
and the anchor rod nut. Plate washers with holes z in.
larger than the anchor rods can be welded to the base plate
so that minimal slip will occur. Alternatively, a setting plate
can be used, and the base plate of the column welded to the
setting plate. The setting plate thickness must be determined
for proper bearing against the anchor rods.
8.2 PARTIAL BASE FIXITY
In some cases, the designer may want to consider design-
ing a column base that is neither pinned nor fixed. These
may be cases where full fixity cannot be obtained or where
the designer wants to know the effect of partial fixity. The
treatment of partial fixity is beyond the scope of this Design
Guide; however, an excellent treatment of partial fixity can
be found in the paper, “Stiffness Design of Column Bases”
(Wald and Jaspart, 1998).
can be neglected because the maximum design wind and
seismic loads occur infrequently. However, for anchor rods
used to anchor machinery or equipment where the full
design loads may occur more often, fatigue should be con-
sidered. In addition, in buildings where crane load cycles are
significant, fatigue should also be considered. AIST TR-13
for the design of steel mill buildings recommends that 50%
of the maximum crane lateral loads or side thrust be used for
fatigue considerations.
In the past, attempts have been made to pretension or pre-
load anchor rods in the concrete to prevent fluctuation of the
tensile stress in anchor rods and, therefore, eliminate fatigue
concerns. This is not recommended unless the anchor rods
are retensioned to accommodate creep in the supporting con-
crete foundation. If setting nuts are employed below the base
plate, pretensioning can be employed to provide a tight con-
nection between the base plate and the anchors.
Fatigue provisions for bolts and threaded parts can be
determined as provided in AISC Specification Appendix 3.
Step 3
Base plate thickness may be governed by bending associated
with compressive loads or tensile loads. However, for lightly
loaded base plates where the dimensions “m” and “n” (as
defined in this procedure) are small, thinner base plate thick-
nesses can be obtained using yield line theory.
For tensile loads, a simple approach is to assume the
anchor rod loads generate bending moments in the base
plate consistent with cantilever action about the web or
flanges of the column section (one-way bending). If the web
is taking the anchor load from the base plate, the web and
its attachment to the base plate should be checked. A more
refined analysis for anchor rods positioned inside the column
flanges would consider bending about both the web and the
column flanges (two-way bending). For the two-way bend-
ing approach, the derived bending moments should be con-
sistent with compatibility requirements for deformations in
the base plate. In either case, the effective bending width for
the base plate can be conservatively approximated using a
45° distribution from the centerline of the anchor rod to the
face of the column flange or web. Calculations for required
base plate thickness for uplift (tensile) loads are illustrated in
AISC Design Guide 1.

Chapter 9
Serviceability Criteria
The design of the lateral load envelope (i.e., the roof bracing
and wall-support system) must provide for the code imposed
loads, which establish the required strength of the structure.
A second category of criteria establishes the serviceability
limits of the design. These limits are rarely codified and
are often selectively applied project by project based on the
experience of the parties involved.
In AISC Design Guide 3, several criteria are given for the
control of building drift and wall deflection. These criteria,
when used, should be presented to the building owner as
they help establish the quality of the completed building.
To be useful, a serviceability criterion must set forth
three items: loading, performance limits, and an analysis
approach. Concerning lateral forces, the loading recom-
mended by AISC Design Guide 3 is the pressure due to
wind speeds associated with a 10-year recurrence interval.
These pressures are approximately 75% of the pressures for
strength design criteria, based on a 50-year return period.
The establishment of deflection limits will follow, with crite-
ria given for each of the wall types previously presented. The
author recommends that frame drift be calculated using the
bare steel frame only. Likewise, the calculations for deflec-
tion of girts should be made using the bare steel section. The
contribution of nonstructural components acting compos-
itely with the structure to limit deflection is often difficult to
quantify. Thus, the direct approach (neglecting nonstructural
contribution) is recommended, and the loads and limits are
calibrated to this analysis approach. The deflection limits for
the various roof and wall systems are discussed in the fol-
lowing sections.
9.1 SERVICEABILITY CRITERIA FOR ROOF
DESIGN
In addition to meeting strength criteria in the design of the
roof structure, it is also necessary to provide for the proper
performance of elements and systems attached to the roof,
such as roofing, ceilings, and hanging equipment. This
requires the control of deflections in the roof structure. Vari-
ous criteria have been published by various organizations.
These limits are:
1. Steel Deck Institute (SDI, 2017b):
a. Maximum deflection of deck due to uniformly dis-
tributed live load: span over 240.
b. Maximum deflection of deck due to a 200-lb con-
centrated load at midspan on a 1-ft section of deck:
span over 240.
2. Steel Joist Institute (SJI, 2015b):
a. Maximum deflection of joists supporting plaster
ceiling due to design live load: span over 360.
b. Maximum deflection of joists supporting ceilings
other than plaster ceilings due to design live load:
span over 240.
3. National Roofing Contractors Association (NRCA,
2015):
a. Maximum deck deflection due to full uniform
load: span over 240.
b. Maximum deck deflection due to a 300-lb load at
midspan: span over 240.
c. Maximum roof structure deflection due to total
load: span over 240.
4. FM Global (2019):
a. Maximum deck deflection due to a 300-lb concen-
trated load at midspan: span over 200.
AISC Design Guide 3 also presents deflection limits for
purlins supporting structural steel roofs (both through fas-
tener types and standing seam types). First, a limiting deflec-
tion of span over 150 for snow loading is recommended.
Secondly, attention is drawn to conditions where a flexible
purlin parallels nonyielding construction, such as at the
building eave. In this case, deflection should be controlled
to maintain positive roof drainage. The appropriate design
load is suggested as dead load plus 50% of snow load or
dead load plus 5-psf live load to check for positive drainage
under load.
Mechanical equipment, hanging conveyors, and other
roof-supported equipment have been found to perform ade-
quately on roofs designed with deflection limits in the range
of span over 150 to span over 240, but these criteria should
be verified with the equipment manufacturer and building
owner. Consideration should also be given to differential
deflections and localized loading conditions.
9.2 METAL WALL PANELS
Relative to serviceability, metal wall panels have two desir-
able attributes—their corrugated profiles make them fairly
limber for out-of-plane distortions and their material and
fastening scheme are ductile (i.e., distortions and possible
yielding do not produce fractures). Also, the material for
edge and corner flashing and trim typically allows move-
ment and distortion without failure. Because of this, the

deflection limits associated with metal panel buildings are
relatively generous. From AISC Design Guide 3 (West et al.,
2003) they are:
1. Frame deflection (drift) perpendicular to the wall sur-
face of frame: eave height divided by 60 to 100.
2. The deflection of girts and wind columns should be
limited to span over 120, unless wall details and wall
supported equipment require stricter limits.
9.3 PRECAST WALL PANELS
Non-load-bearing precast wall panels frequently span from
grade to eave as simple-span members. Therefore, drift does
not change the statics of the panel. The limitation on drift in
the building frame is established to control the amount of
movement in the joint at the base of the panel as the frame
drifts. This limit has been proposed to be eave height over
100.A special case exists when precast panels are set atop the
perimeter foundations to eliminate a grade wall. The foun-
dation anchorage, the embedment of the panel in the soil,
and the potential of the floor slab to act as a fulcrum mean
that the frame deflections must be analyzed for compatibility
with the panel design. It is possible to tune frame drift with
panel stresses, but this requires interaction between frame
designer and panel designer. Usually the design of the frame
precedes that of the panel. In this case, the frame behavior
and panel design criteria should be carefully specified in the
construction documents.
9.4 MASONRY WALLS
Masonry walls may be hollow, grouted, solid, or grouted
and reinforced. Masonry itself is a brittle, nonductile mate-
rial. Masonry with steel reinforcement has ductile behavior
overall but will show evidence of cracking when subjected
to loads that stress the masonry in tension. When masonry
is attached to supporting steel framework, deflection of the
supports may induce stresses in the masonry. It is rarely
feasible to provide sufficient steel (stiffness) to keep the
masonry stresses below cracking levels. Thus, flexural
tension cracking in the masonry is likely and, when properly
detailed, is not considered a detriment. The correct strategy
is to impose reasonable limits on the support movements and
detail the masonry to minimize the impact of cracking.
Masonry should be provided with vertical control joints at
the building columns and wind columns. This prevents flex-
ural stresses on the exterior face of the wall at these locations
from inward wind. Because the top of the wall is generally
free to rotate, no special provisions are required there. Most
difficult to address is the base of the wall joint. To carry the
weight of the wall, the base joint must be solid, not caulked.
Likewise, the mortar in the joints makes the base of the wall
a fixed condition until the wall cracks.
Frame drift recommendations are set to limit the size of
the inevitable crack at the base of the wall. Because rein-
forced walls can spread the horizontal base cracks over sev-
eral joints, separate criteria are given for them. If proper base
joints are provided, reinforced walls can be considered as
having the behavior of precast walls—that is, simple-span
elements with pinned bases. In that case, the limit for precast
wall panels would be applicable. Where wainscot walls are
used, consideration must be given to the joint between metal
wall panel and masonry wainscot. The relative movements
of the two systems in response to wind must be controlled to
maintain the integrity of the joint between the two materials.
The recommended limits for the deflection of elements
supporting masonry are (West et al., 2003):
1. Frame deflection (drift) perpendicular to an unrein-
forced wall should allow no more than a z-in. crack
to open in one joint at the base of the wall. The drift
allowed by this criterion can be conservatively calcu-
lated by relating the wall thickness to the eave height
and taking the crack width at the wall face as z-in.
and zero at the opposite face.
2. Frame deflection (drift) perpendicular to a reinforced
wall is recommended to be eave height over 100.
3. The deflection of wind columns and girts should be
limited to span over 240, but not greater than 1.5 in.

PART 2
INDUSTRIAL BUILDINGS WITH CRANES
Chapter 10
Introduction to Part 2
This section of the Guide deals with crane buildings and
will include coverage of those aspects of industrial build-
ings peculiar to the existence of overhead and underhung
cranes. In that context, the major difference between crane
buildings and other industrial buildings is the frequency of
loading caused by the cranes. Thus, crane buildings should
be classified for design purposes according to the frequency
of loading.
Crane building classifications have been established in the
Guide for the Design and Construction of Mill Buildings,
AIST TR-13 (AIST, 2003),as ClassesA, B, C, and D. Classi-
fications for cranes have been established by the Crane Man-
ufacturers Association of America (CMAA), Specification
for Top Running Bridge and Gantry Type Multiple Girder
Electric Overhead Traveling Cranes—No. 70 (CMAA,
2015a), hereafter referred to as CMAA 70. These designa-
tions should not be confused with the building designations.
10.1 AIST TR-13 BUILDING CLASSIFICATIONS
Class A is a building in which members may experience
either 500,000 to 2,000,000 repetitions or over 2,000,000
repetitions in the estimated life span of the building of
approximately 50 years. The owner must analyze the service
and determine which load condition may apply. It is recom-
mended that the following building types be considered as
Class A:
Batch annealing buildings
Scrap yards
Billet yards
Skull breakers
Continuous casting buildings
Slab yards
Foundries
Soaking pit buildings
Mixer building
Steelmaking buildings
Mold conditioning buildings
Stripper buildings
Scarfing yards
Other buildings based on predicted operational
requirements
Class B is a building in which members may experience
a repetition of 100,000 to 500,000 cycles of a specific load-
ing or 5 to 25 repetitions of such load per day for a life of
approximately 50 years.
Class C is a building in which members may experience a
repetition of 20,000 to 100,000 cycles of a specific loading
during the expected life of a structure, or one to five rep-
etitions of such load per day for a life of approximately 50
years.
Class D is a building in which no member will experience
more than 20,000 repetitions of a specific loading during the
expected life of a structure.
10.2 CMAA CRANE CLASSIFICATIONS
The following classifications are taken directly from
CMAA 70.
10-2 CRANE CLASSIFICATIONS
2.1 Service classes have been established so that the most
economical crane for the installation may be specified in
accordance with this specification.
The crane service classification is based on the load
spectrum reflecting the actual service conditions as
closely as possible.
Load spectrum is a mean effective load, which is uni-
formly distributed over a probability scale and applied to
the equipment at a specified frequency. The selection of
the properly sized crane component to perform a given
function is determined by the varying load magnitudes
and given load cycles which can be expressed in terms of
the mean effective load factor.
K W P W P W P
n n
1
3
1 2
3
2
3
3
= + W P
3
3
3
+ + (10-1)
where
K =
Mean effective load factor (used to establish crane
service class only).
P =
Load probability; expressed as a ratio of cycles
under each load magnitude condition to the total

cycles. The sum total of the load probabilities P
must equal 1.0.
W =
Load magnitude; expressed as a ratio of each lifted
load to the rated capacity. Operation with no lifted
load and the weight of any attachment must be
included.
All classes of cranes are affected by the operating con-
ditions, therefore for the purpose of the classifications,
it is assumed that the crane will be operating in normal
ambient temperature of 0° to 104°F (−17.7° to 40°C) and
normal atmospheric conditions (free from excessive dust,
moisture and corrosive fumes).
The cranes can be classified into loading groups accord-
ing to the service conditions of the most severely loaded
part of the crane. The individual parts that are clearly sep-
arate from the rest or forming a self-contained structural
unit can be classified into different loading groups if the
service conditions are fully known.
2.2 CLASS A (STANDBY OR INFREQUENT
SERVICE)
This service class covers cranes which may be used in
installations such as powerhouses, public utilities, turbine
rooms, motor rooms and transformer stations where pre-
cise handling of equipment at slow speeds with long, idle
period between lifts are required. Capacity loads may be
handled for initial installation of equipment and for infre-
quent maintenance.
2.3 CLASS B (LIGHTSERVICE)
This service covers cranes which may be used in repair
shops, light assembly operations, service buildings, light
warehousing, etc., where service requirements are light
and the speed is slow. Loads may vary from no load to
occasional full rated loads with two to five lifts per hour,
averaging ten feet per lift.
2.4 CLASS C (MODERATE SERVICE)
This service covers cranes which may be used in machine
shops or paper mill machine rooms, etc., where service
requirements are moderate. In this type of service the
crane will handle loads which average 50% of the rated
capacity with 5 to 10 lifts per hour, averaging 15 ft, not
over 50% of the lift at rated capacity.
2.5 CLASS D (HEAVY SERVICE)
This service covers cranes which may be used in heavy
machine shops, foundries, fabricating plants, steel ware-
houses, container yards, lumber mills, etc., and standard
duty bucket and magnet operations where heavy duty pro-
duction is required. In this type of service, loads approach-
ing 50% of the rated capacity will be handled constantly
during the working period. High speeds are desirable for
this type of service with 10 to 20 lifts per hour averaging
15 ft, not over 65% of the lifts at rated capacity.
2.6 CLASS E (SEVERE SERVICE)
This type of service requires a crane capable of handling
loads approaching a rated capacity throughout its life.
Applications may include magnet/bucket combination
cranes for scrap yards, cement mills, lumber mills, fertil-
izer plants, container handling, etc., with twenty or more
lifts per hour at or near the rated capacity.
2.7 CLASS F (CONTINUOUS SEVERE SERVICE)
This type of service requires a crane capable of handling
loads approaching rated capacity continuously under
severe service conditions throughout its life. Applications
may include custom designed specialty cranes essential
to performing the critical work tasks affecting the total
production facility. These cranes must provide the highest
reliability with special attention to ease of maintenance
features.
The class of crane, the type of crane, and loadings all
affect the design. The fatigue associated with crane class is
especially critical for the design of crane runways and con-
nections of crane runway beams to columns. Classes E and
F produce particularly severe fatigue conditions. The deter-
mination of fatigue stress levels and load conditions is dis-
cussed in more detail in the next section.
The CMAA 70 crane classifications do not relate directly
to the AISC Specification Appendix 3 fatigue provisions.
Fatigue provisions in previous versions of the AISC Speci-
fication, such as the 1989 edition (AISC, 1989), included
“Loading Conditions” based on the number of loading
cycles. These loading conditions, while not directly related
to the CMAA crane classifications, were presented in the
previous editions of this Design Guide corresponding to the
average number of lifts for each CMAA 70 crane classifica-
tion. They are included here in Table 10-1 because they may
be of value in some instances.
The approximate number of loading cycles for each
loading condition is given in the 1989 AISC Specification
Table A-K4.1. The table is repeated here as Table 10-2.
The AISC Specification no longer refers to loading condi-
tions and provides equations to determine an allowable stress
range for a given number of stress cycles. The AISC Speci-
fication states that, “The engineer of record shall provide
either complete details including weld sizes or shall specify
the planned cycle life and the maximum range of moments,
shears and reactions for the connections.” To use the equa-
tions, the designer must enter the value of nSR, which is the
stress range fluctuation in design life, into the appropriate
design equations provided in AISC Specification Appen-
dix 3. If required, nSR can be estimated from Table 11-1.

The 2016 AISC Specification fatigue provisions are the
most up-to-date provisions and are recommended for use by
the author.
The designer must carefully consider whether or not the
requirements of AIST TR-13 must be followed for a given
structure or whether only the requirements of the building
code need be followed. This decision is generally made
based on the CMAA crane classifications and the AIST
TR-13 building classifications. For CMAA Crane Classifi-
cations A, B, C, and in some cases Class D, the AIST TR-13
requirements are generally not required. These differences
should be kept in mind by the reader of this Design Guide.
Table 10-1. Crane Loading Conditions
CMAA Crane Classification 1989 AISC Specification Loading Condition
A, B 1
C, D 2
E 3
F 4
Table 10-2. AISC Loading Cycles
Loading Condition From To
1 20,000a
100,000b
2 100,000 500,000c
3 500,000 2,000,000d
4 Over 2,000,000
a
Approximately equivalent to two applications every day for 25 years.
b
Approximately equivalent to 10 applications every day for 25 years.
c
d

Chapter 11
Fatigue
Proper functioning of bridge cranes is dependent upon
proper crane runway girder design and detailing. The run-
way design must account for the fatigue effects caused by
the repeated passing of the crane. Runway girders should
be thought of as a part of a system comprised of the crane
rails, rail attachments, electrification support, crane stops,
crane column attachment, tie backs, and the girder itself. All
of these items should be incorporated into the design and
detailing of the crane runway girder system.
Crane runway girder problems are associated with fatigue
cracking in the majority of situations. However, engineers
have designed crane runway girders that have performed
with minimal problems while being subjected to millions of
cycles of loading. The girders that are performing success-
fully have been properly designed and detailed to:
• Limit the applied stress range to acceptable levels.
• Avoid unexpected restraints at the attachments and
supports.
• Avoid stress concentrations at critical locations.
• Avoid eccentricities due to rail misalignment or crane
travel and other out-of-plane distortions.
• Minimize residual stresses.
Even when all state-of-the-art design provisions are fol-
lowed, building owners can expect to perform periodic
maintenance on runway systems. Runway systems that have
performed well have been properly maintained by keeping
the rails and girders aligned and level.
Some fatigue damage should be anticipated eventually,
even in well-designed structures, because fabrication and
erection cannot be perfect. Fabrication, erection, and main-
tenance of tolerance requirements in the AISC Code of
Standard Practice, the American Welding Society (AWS)
Structural Welding Code—Steel (AWS, 2015), and AIST
TR-13 should be followed in order to provide predicted
fatigue behavior.
Fatigue provisions have a 95% reliability factor (two stan-
dard deviations below mean curve of test data) for a given
stress range and expected life condition. Thus, it is reason-
able to expect that 5% of similar details can experience
fatigue failure before the expected fatigue life is expired.
However, if the designer chooses a design life of the struc-
ture to be shorter than the expected fatigue life according
to AISC criteria, the reliability of a critical detail should be
higher than 95%.
11.1 FATIGUE DAMAGE
Fatigue damage can be characterized as progressive crack
growth due to fluctuating stress on the member. Fatigue
cracks initiate at small defects or imperfections in the base
material or weld metal. The imperfections act as stress risers
that magnify the applied elastic stresses into small regions
of plastic stress. As load cycles are applied, the plastic
strain in the small plastic region advances until the material
separates and the crack advances. At that point, the plastic
stress region moves to the new tip of the crack, and the pro-
cess repeats itself. Eventually the crack size becomes large
enough that the combined effect of the crack size and the
applied stress exceed the toughness of the material, and a
final fracture occurs.
Fatigue failures result from repeated application of service
loads, which cause crack initiation and propagation to final
fracture. The dominant variable is the tensile stress range
imposed by the repeated application of the live load, not
the maximum stress that is imposed by live plus dead load.
Fatigue damage develops in three stages: crack initiation,
stable crack growth, and unstable crack growth to fracture.
Of these, the crack initiation phase takes up about 80% of
the total fatigue life; thus, when cracks are of detectible size,
the fatigue life of a member or detail is virtually exhausted,
and prompt remedial action should be taken. Abrupt changes
in cross section, geometrical discontinuities such as toes of
welds, unintentional discontinuities from lack of perfection
in fabrication, and effects of corrosion and residual stresses
all have a bearing on the localized range of tensile stress
in details that lead to crack initiation. These facts make it
convenient and desirable to structure fatigue design provi-
sions on the basis of categories that reflect the increase in
tensile stress range due to the severity of the discontinuities
introduced by typical details. Application of stress concen-
tration factors to stresses determined by typical analysis is
not appropriate.
However, fluctuating compressive stresses in a region of
tensile residual stress may cause a net fluctuating tensile
stress or reversal of stress that may cause cracks to initiate.
The AISC Specification provides continuous functions in
terms of cycles of life and stress range in lieu of the previ-
ous criteria for fatigue life that reflected the database only at
the break points in the step-wise format. AISC Specification
Appendix 3 uses a single table, Table A-3.1, that is divided
into sections that describe various conditions. The sections
are:

1. Plain material away from any welding
2. Connected material in mechanically fastened joints
3. Welded joints joining components of built-up
members
4. Longitudinal fillet-welded end conditions
5. Welded joints transverse to direction of stress
6. Base metal at welded transverse member connections
7. Base metal at short attachments
8. Miscellaneous
AISC Specification Appendix 3 uses equations to calcu-
late the allowable stress range for a selected design life for
various conditions and associated stress categories.
The point of potential crack initiation is identified by
description and shown in figures in Table A-3.1. The tables
contain the threshold design stress, FTH, for each stress cat-
egory, and also provide the detail constant, Cf, applicable to
the stress category that is required for calculating the design
stress range, FSR. For the majority of stress categories:
F
C
n
F
1,000
SR
f
SR
TH
0.333
=
⎛
⎝
⎜
⎞
⎠
⎟ ≥
(Spec. Eq. A-3-1)
where
Cf =
constant taken from AISC Specification Table
A-3.1 for the applicable stress category
FSR = allowable stress range, ksi
FTH =
threshold allowable stress range, maximum stress
range for indefinite design life from AISC Specifi-
cation Table A-3.1, ksi
nSR = number of stress range fluctuations in design life
The 2016 AISC Specification, as well as previous editions
of the AISC Specification, limits the allowable stress range
for a given service life based on an anticipated severity of the
stress riser for a given fabricated condition.
CMAA 70 includes crane designations that define the
anticipated number of full uniform amplitude load cycles for
the life of the crane. The correlation of the CMAA crane
designations for a given crane to the required fatigue life for
the structure cannot be directly determined. The crane does
not lift its maximum load or travel at the same speed every
day or every hour. Shown in Table 11-1 are estimates of the
number of cycles of full uniform amplitude for CMAA crane
classifications A through F over a 40-year period. It must be
emphasized that these are only guidelines, and actual duty
cycles can only be established from the building’s owner and
the crane manufacturer.
Consideration of fatigue requires that the designer deter-
mine the anticipated number of full uniform amplitude load
cycles. To properly apply the AISC Specification fatigue
equations to crane runway girder fatigue analyses, one must
understand the difference between the AISC fatigue provi-
sions determined using data from cyclic constant amplitude
loading tests and crane runway variable amplitude cyclic
loadings. When AIST TR-13 is not specified, it is common
practice for the crane runway girder to be designed for a ser-
vice life that is consistent with the crane classification.
It should be noted that one crane will often involve more
than one cycle on the structure (e.g., one cycle fully loaded
and one cycle from the unloaded crane). In the evaluation of
some runways, it is important to determine the cumulative
fatigue damage resulting from variable amplitude loading.
The reader is referred to the Palmgren-Miner rule (Collins,
1981). Also, for some details the passage of each wheel
on the crane may be critical as compared to the total crane
passage.
11.2 CRANE RUNWAY FATIGUE
CONSIDERATIONS
The AISC Specification provisions as they relate to crane
runway design will follow. A complete design example is
provided in Fisher and Van de Pas (2002). The fatigue provi-
sions discussed in the following assume that the girders are
fabricated using the AWS provisions for cyclically loaded
structures. In a few instances, additional weld requirements
are recommended by AIST TR-13. These are pointed out in
the following sections.
Table 11-1. CMAA Classification vs. Design Life
CMAA Crane Classification Design Life
A 20,000
B 50,000
C 100,000
D 500,000
E 1,500,000
F 2,000,000

Tension Flange Stress
When runway girders are fabricated from plate material,
fatigue requirements are more severe than for rolled shape
girders. AISC Specification Appendix 3, Table A-3.1, Sec-
tion 1, applies to plain material. Appendix 3, Table A-3.1,
Section 3.3, applies to welded joints joining components of
built-up members. Stress Category B is required for plate
girders as compared to stress Category A for rolled shapes.
Web-to-Flange Welds
Appendix 3, Table A-3.1, Section 8.2, applies to shear in
fillet welds that connect the web to tension and compres-
sion flanges, which is stress Category F. Cracks have been
observed in plate girders at the junction of the web to the
compression flange of runway girders when fillet welds are
used to connect the web to the compression flange. Such
cracking has been traced to localized tensile bending stresses
in the bottom side of the compression flange plate with each
wheel load passage. Each wheel passage may occur two or
four or more times with each passage of the crane; thus, the
life cycles for this consideration is generally several times
greater than the life cycles to be considered in the girder live
load stress ranges due to passage of the loaded crane. The
calculation of such highly localized tensile bending stresses
is so complex and unreliable that the problem is buried in
conservative detail requirements. To reduce the likelihood of
such cracks,AIST TR-13 recommends that the top flange-to-
web joint be a CJP weld with fillet reinforcement.
Tiebacks
Tiebacks are provided at the end of the crane runway girders
to transfer lateral forces from the girder top flange into the
crane column and to laterally restrain the top flange of the
crane girder against buckling; see Figure 11-1. The tiebacks
must have adequate strength to transfer the lateral crane
loads. However, the tiebacks must also be flexible enough
to allow for longitudinal movement of the top of the girder
caused by girder-end rotation. The amount of longitudinal
movement due to the end rotation of the girder can be sig-
nificant. The end rotation of a 40-ft girder that has undergone
a deflection of l/600 is about 0.005 rad. For a 36-in.-deep
Fig. 11-1. Tieback detail.

girder, this results in 0.2 in. of horizontal movement at the
top flange. The tieback must also allow for vertical move-
ment due to axial shortening of the crane column. This ver-
tical movement can be in the range of 4 in. In general, the
tieback should be attached directly to the top flange of the
girder. Attachment to the web of the girder with a diaphragm
plate should be avoided. The lateral load path for this detail
causes bending stresses in the girder web perpendicular to
the girder cross section. The diaphragm plate also tends to
resist movement due to the axial shortening of the crane col-
umn. Various AISC fatigue provisions are applicable to the
loads depending on the exact tieback configurations. In addi-
tion, a variety of proprietary tieback solutions designed to
provide the required flexibility and lateral load resistance are
available from crane runway product suppliers.
Bearing Stiffeners
Bearing stiffeners should be provided at the ends of the gird-
ers as required by the AISC Specification. Fatigue cracks
have occurred at the connection between the bearing stiff-
ener and the girder top flange. The cracks occurred in details
where the bearing stiffener was fillet welded to the underside
of the top flange. Passage of each crane wheel produces shear
stress in the fillet welds. The AISC Specification fatigue
provisions contain fatigue criteria for fillet welds in shear;
however, the determination of the actual stress state in the
welds is extremely complex; thus, AIST TR-13 recommends
that CJP welds be used to connect the top of the bearing
stiffeners to the top flange of the girder. The bottom of the
bearing stiffeners may be fitted (preferred) or fillet welded to
the bottom flange. All stiffener-to-girder web welds should
be continuous. Horizontal cracks have been observed in the
webs of crane girders with partial-height bearing stiffeners.
The cracks start between the bearing stiffeners and the top
flange and run longitudinally along the web of the girder.
There are many possible causes for the propagation of these
cracks. One possible explanation is that eccentricity in the
placement of the rail on the girder causes distortion and rota-
tion of the girder cross section.
Intermediate Stiffeners
If intermediate stiffeners are used, AIST TR-13 also recom-
mends that the intermediate stiffeners be welded to the top
flange with CJP welds for the same reasons as with bear-
ing stiffeners. Stiffeners are permitted to be stopped short of
the tension flange in accordance with the AISC Specification
Section G2.3 provisions.AIST TR-13 also recommends con-
tinuous stiffener-to-web welds for intermediate stiffeners.
Fatigue must be checked where the stiffener terminates
adjacent to the tension flange. This condition is addressed
in AISC Specification Appendix 3, Table A-3.1, Section 5.7.
Cap Channels and Cap Plates
Cap channels or cap plates are frequently used to provide
adequate top flange capacity to transfer lateral loads to the
crane columns and to provide lateral torsional stability of the
runway girder cross section. It should be noted that the cap
channel or plate does not fit perfectly with 100% bearing on
the top of the wide flange. The tolerances given in ASTM A6
allow the wide-flange member to have some flange tilt along
its length, the plate may be cupped or slightly warped, or the
channel may have some twist along its length. These con-
ditions will leave small gaps between the top flange of the
girder and the top plate or channel. The passage of the crane
wheel over these gaps will tend to distress the channel or
plate to top flange welds. Calculation of the stress condition
for these welds is not practical. Because of this phenome-
non, cap plates or channels should not be used with Class E
or F cranes. For less severe duty cycle cranes, shear flow
stress in the welds can be calculated and limited according
to the AISC Specification fatigue provisions in Appendix 3,
Table A-3.1, Section 8.2. The channel or plate welds to the
top flange can be continuous or intermittent. However, the
AISC design stress range for the base metal is reduced from
Stress Category B (Appendix 3, Table A-3.1, Section 3.1)
for continuous welds to Stress Category E (Appendix 3,
Table A-3.1, Section 3.4) for intermittent welds. Because of
the reduced fatigue life for intermittent welds and the poten-
tial stresses caused by a gap under the cap channel, it is rec-
ommended that only continuous welds be used to connect
the cap channel to the beam flange.
To eliminate the fatigue problems associated with cap
channel welds, angles can be welded to the top flange of the
runway girders to provide the required lateral stability and to
transfer the lateral loads.
Crane Column Cap Plates
The crane column cap plate should be detailed so it does not
restrain the end rotation of the girder. If the cap plate girder
bolts are placed between the column flanges, the girder end
rotation is resisted by a force couple between the column
flange and the bolts. This detail has been known to cause bolt
failures. Preferably, the girder should be bolted to the cap
plate outside of the column flanges. The column cap plate
should be extended outside of the column flange with the
bolts to the girder placed outside of the column flanges. The
column cap plate should not be made overly thick, as this
detail requires the cap plate to distort to allow for the end
rotation of the girder. The girder-to-cap-plate bolts should be
adequate to transfer the tractive or bumper forces to the lon-
gitudinal crane bracing. Traction plates between girder webs
may be required for large tractive forces or bumper forces.

The engineer should consider using finger tight bolts with
upset threads as a means of reducing bolt fatigue in crane
column cap plates (Rolfes and Fisher, 2001).
Miscellaneous Attachments
Attachments to crane runway girders should be avoided.
AIST TR-13 specifically prohibits welding attachments to
the tension flange of runway girders. Brackets to support the
runway electrification are often necessary. If the brackets
are bolted to the web of the girder, fatigue consequences are
relatively minor. However, if the attachment is made with
fillet welds to the web, AISC Specification Appendix 3,
Table A-3.1, Section 7.2, applies. This provision places the
detail into Stress Category D or E, depending on the detail.
If transverse stiffeners are present, the brackets could be
attached to the stiffeners. The girders should be detailed so
that the bracket holes are shown to eliminate the need for
field drilling.

Chapter 12
Crane-Induced Loads and Load Combinations
It is recommended that the designer show the crane wheel
loads, wheel spacing, bumper forces, and the design criteria
used to design the structure on the drawings.
Although loading conditions for gravity, wind, and seis-
mic loads are well defined among building codes and stan-
dards, crane loading conditions generally are not.
As mentioned previously, crane fatigue loads are primar-
ily a function of the class of service, which in turn is based
primarily on the number of cycles of a specific loading case.
This classification should be based on the estimated life
span, rate of loading, and the number of load repetitions.
The owner should specify or approve the classification for
all portions of a building. A maximum life span of 50 years
is typically acceptable.
The provisions of ASCE/SEI 7-16 and AIST TR-13 for
crane runway loads are summarized in the following discus-
sion. ASCE/SEI 7-16 is referenced by the IBC and is a legal
requirement. AIST TR-13 is a guideline and can be used for
situations not covered by ASCE/SEI 7-16, or when speci-
fied by project specifications. In addition, the Metal Build-
ing Manufacturers Association (MBMA) Low-Rise Building
Systems Manual (MBMA, 2012) provides a comprehensive
discussion on crane loads.
For LRFD load combinations, ASCE/SEI 7-16 indicates
that the live load of a crane is the rated capacity, thus a 1.6
load factor is to be used. No comments are made about
appropriate load factors relative to the trolley, hoist, or
bridge weight. The author recommends using a 1.2 load fac-
tor for the bridge weight and the hoist and trolley weight.
12.1 VERTICAL IMPACT
ASCE/SEI 7-16
ASCE/SEI 7-16 defines the maximum wheel load as “The
maximum wheel loads shall be the wheel loads produced by
the weight of the bridge, as applicable, plus the sum of the
rated capacity and the weight of the trolley with the trolley
positioned on its runway at the location where the result-
ing load effect is maximum.” Vertical impact percentages are
then multiplied by the maximum wheel loads. The percent-
age factors contained in ASCE/SEI 7-16 are as follows:
Monorail cranes (powered) 25%
Cab-operated or remotely operated bridge cranes
(powered)
25%
Pendant-operated bridge cranes (powered) 10%
Bridge cranes or monorail cranes with hand-geared
bridge, trolley, and hoist
0
AIST TR-13
The allowances for vertical impact are specified as 25% of
the maximum wheel loads for all crane types, except a 20%
impact factor is recommended for motor room maintenance
cranes.
In all cases, impact loading should be considered in the
design of column brackets regardless of whether ASCE/SEI
7-16 or AIST TR-13 requirements are being used.
12.2 SIDE THRUST
Horizontal forces act on crane runways due to a number of
factors including:
1. Runway misalignment
2. Crane skew
3. Trolley acceleration
4. Trolley braking
5. Crane steering
ASCE/SEI 7-16
“The lateral force on crane runway beams with electrically
powered trolleys shall be calculated as 20% of the sum of
the rated capacity of the crane and the weight of the hoist
and trolley. The lateral force shall be assumed to act hori-
zontally at the traction surface on a runway beam, in either
direction perpendicular to the beam, and shall be distributed
with due regard to the lateral stiffness of the runway beam
and supporting structure.” ASCE/SEI 7-16 is silent on what
load factor to use with the lateral crane forces. The author
recommends using a 1.6 factor on the lateral thrust because
these loads are not known with the same degree of accuracy
as the bridge and hoist weights.
AIST TR-13
AIST TR-13 requires that “The recommended total side
thrust shall be distributed with due regard for the lateral stiff-
ness of the structures supporting the rails and shall be the
greatest of:
1. That specified in Table 3.2 (shown here as Table 12-1).
2. 20% of the combined weight of the lifted load and
trolley. For stacker cranes this factor shall be 40% of
the combined weight of the lifted load, trolley, and
rigid arm.
3. 10% of the combined weight of the lifted load and

crane weight. For stacker cranes this factor shall be
15% of the combined weight of the lifted load and the
crane weight.”
In AIST TR-13, lifted load is defined as “a total weight
lifted by the hoist mechanism, including working load,
all hooks, lifting beams, magnets or other appurtenances
required by the service but excluding the weight of column,
ram or other material handling device that is rigidly guided
in a vertical direction during hoisting action.” For pendant
operated cranes, the AIST TR-13 side thrust is taken as 20%
of the maximum load on the driving wheels. In most cases,
half of the wheels are driving wheels. AIST TR-13 requires
that radio-operated cranes be considered as cab-operated
cranes with regard to side thrusts.
Table 12-2 is provided to illustrate the variation between
ASCE/SEI 7-16 and AIST TR-13 for a particular crane size.
12.3 LONGITUDINAL OR TRACTIVE FORCE
ASCE/SEI 7-16
The longitudinal force on crane runway beams is calculated
as 10% of the maximum wheel loads of the crane.ASCE/SEI
7-16 excludes bridge cranes with hand-geared bridges from
this requirement; thus, the author presumes that it is ASCE’s
Table 12-1. AIST TR-13 Crane Side Thrusts (AIST, 2003)
Crane Type Total Side Thrust, % of Cap Lifted Load
Mill cranes 40
Ladle cranes 40
Clamshell bucket and magnet cranes
(including slab and billet yard cranes)
100
Soaking pit cranes 100
Stripping cranes 100a
Motor room maintenance cranes 30
Stacker cranes (cab-operated) 200
a
ingot and mold
Table 12-2. ASCE/SEI-7 vs. AIST TR-13 Side Thrust Comparison
100T Mill Crane, Trolley Weight = 60,000 lb (Includes Hoist), Entire Crane Weight = 157,000 lb
Criteria Equation (Total Force) Total Force, kips
ASCE/SEI-7 (ASD) 0.20 (Rated capacity + trolley weight) 52.0
AIST TR-13
(1) 0.40 (lifted load)
(2) 0.20 (lifted load + trolley weight)
(3) 0.10 (lifted load + entire crane weight)
80.0
52.0
35.7
position that tractive forces are not required for hand-geared
bridge cranes. It is further stated in ASCE/SEI 7-16 that the
longitudinal force shall be assumed to act horizontally at the
traction surface of a runway beam in either direction parallel
to the beam. ASCE is silent on the LRFD load factor for the
longitudinal force. The author recommends using 1.6.
AIST TR-13
The tractive force is taken as 20% of the maximum load on
driving wheels.
12.4 CRANE STOP FORCES
The magnitude of the bumper force is dependent on the
energy absorbing device used in the crane bumper. The
device may be linear such as a coil spring or nonlinear such
as hydraulic bumpers. See Section 14.6 for additional infor-
mation on the design of the runway stop.
The crane stop, crane bracing, and all members and their
connections that transfer the bumper force to the ground
should be designed for the bumper force. It is recommended
that the designer indicate on the structural drawings the
magnitude of the bumper force assumed in the design. The
bumper force is typically specified by the owner or crane
supplier.

12.5 ECCENTRICITIES
The bending of the column due to eccentricity of the crane
girder on the column seat must be investigated. The critical
bending for this case may occur when the crane is not cen-
tered over the column but located just to one side as illus-
trated in Figure 12-1. Eccentricity of the rail to the girder
centerline should also be considered in the design. Under
this condition, the vertical wheel loads induce torsion on the
girder that is typically resolved into a force couple on the top
and bottom flanges for design. See Chapter 15 for suggested
tolerances for alignment of rail and girder centerlines.
12.6 SEISMIC LOADS
Although cranes do not induce seismic loads on a struc-
ture, the crane weight should be considered in seismic load
determination. The seismic mass of cranes and trolleys that
lift a suspended load need include only the empty weight of
the equipment. The designer should carefully evaluate the
location of the cranes within the building in the seismic anal-
ysis. Where appropriate, a site investigation should be per-
formed in order to determine the soil profile type for seismic
response. Seismic response interaction between the crane
and the building should be taken into account.
Special consideration should also be given to design
requirements beyond those specified in the building code for
buildings, structures, and equipment that must remain ser-
viceable immediately after a design level earthquake. This
may include the examination of vertical accelerations and
their effect on the crane’s ability to not “bounce” off the run-
way during a seismic event.Also, the designer is cautioned to
verify seismic limitations that may be imposed on the struc-
tural system, and to determine the need for special detailing
requirements based on the Seismic Design Category.
12.7 LOAD COMBINATIONS
In addition to the applicable building code, the owner may
require conformance with AIST TR-13. However, if AIST
TR-13 is not specified, the designer should consider the use of
the structure in determining the criteria for the design. Build-
ing codes may not contain information on how to combine
the various crane loads—that is, which crane loads and how
many cranes should be considered loaded at one time—but
typically they do address how crane loads should be combined
with wind, snow, live, seismic, and other loads.
For one crane, each span must be designed for the most
severe loading conditions with the crane in the worst posi-
tion for each element that is affected. As mentioned, when
more than one crane is involved in making a lift, most codes
are silent on a defined procedure. Engineering judgment on
the specific application must be used.
The author recommends the following provisions for the
design of members subject to crane lifts. These provisions
are applicable to the design of supporting elements and are
consistent with those currently proposed for the next edition
of AIST TR-13, which is expected to be published in 2020.
3.10.2.1. LRFD Load Combinations
1. D
1.4
1a. D C
1.4 1.4 dm
+
2. D L L S R
1.2 1.6 0.5 or or
r
( )
+ +
2a. D C L C C C
L S R
1.2 1.6 1.0
0.5 or or
dm vm ss ls
r
( ) ( )
( )
+ + + + + +
2b. D C C C C L
L S R
1.2 1.0
0.5 or or
dm vm ss ls
r
( ) ( )
( )
+ + + + + +
Fig.12-1. Possible critical crane location.

2c. D C C C C L
L S R
1.2 1.6
0.5 or or
ds vs l ls
r
( ) ( )
( )
+ + + + + +
3. D L S R L W
1.2 1.6 or or or 0.5
r
( ) ( )
+ +
3a. D C L S
C
R
C C L W
1.2 1.6 or or
1.0 or 0.5
dm r
vm ss ls
( )
( )
( )
( )
+ + +
+ + +
4. D W L L S R
1.2 1.0 0.5 or or
r
( )
+ + +
4a. D C W L C
L S R
1.2 1.2 1.0
0.5 or or
dm vm
r
( )
+ + + + +
5. D E L S
1.2 1.0 0.2
+ + +
5a. D C E C L S
1.2 1.2 1.0 0.2
dm vm
+ + + + +
6. D W
0.9 1.0
+
7. D E
0.9 1.0
+
7a. D C C E
0.9 or 1.0
dm ds
( )
+ +
8. D C C C
1.2 1.2 1.0 1.0
ds vs bs
+ + +
9. D C C C
0.9 1.6 1.6
ds vs ss
( )
+ + +
(min)
10. D C C C
0.9 1.6 or 1.0
ds ls b
( )
+ +
3.10.2.2. ASD Load Combinations
1. D
1a. D Cdm
+
2. D L
+
2a. D C C C C L
dm vm ss ls
+ + + + +
2b. D C C C C L
ds vs ls i
+ + + + +
3. D L S R
or or
r
( )
+
3a. D C L S R
or or
dm r
( )
+ +
4. D L L S R
0.75 or or
r
[ ]
( )
+ +
4a. D C C C C L
L S R
0.75
or or
dm vm ss ls
r
[
]
( )
+ + + + + +
4b. D C C C C L
L S R
0.75
or or
ds vs i ls
r
[
]
( )
+ + + + + +
5. D W E
0.6 or 0.7
( )
+
5a. D C W E
0.6 or 0.7
dm ( )
+ +
6a1. D L W L S R
0.75 0.6 or or
r
[ ]
( )
+ + +
6a2. D C C L W
L S R
0.75 0.6
or or
dm vm
r
[
]
( )
+ + + + +
6a3. D C C Css + Cls + 0.3W +
L S R
0.75
or or
dm vm
r
[
]
( )
+ + +
6b1. D L E S
0.75 0.7
[ ]
+ + +
6b2. D C C L E S
0.75 0.7
dm vm
[ ]
+ + + + +
7. D W
0.6 0.6
+
8. D C C E
0.6 or 0.7
dm ds
( )
+ +
9. D C C C
0.67
ds vs bs
+ + +
10. D C C C
0.6 ds vs ss
( )
+ + +
( )
min
11. D C C C C
0.6 or 0.67
ds vs ls bs
( ) ( )
+ + +
3.10.2.3. Fatigue
For purposes of fatigue design, crane loads to be consid-
ered are (Cds+Cvs+0.5Css). The number of cycles used as
the basis for fatigue design shall be consistent with the
building classification covered in Section 1.4. The owner
shall designate an increase in the estimated number of
load repetitions for any portion of the building structure
for which the projected work load or possible changes in
building usage warrants.
The variables as defined inAIST TR-13, Section 3.10.1.
are:
C =
crane load (see specific crane load identifications
in the next paragraph)
D = dead load
E = earthquake load
L = live load
Lr = roof live load
R = rain load
S = snow load
W = wind load
Individual crane loads referenced in the load combina-
tions in the preceding text are as follows:
Cbs = crane bumper force
Cds =
crane dead load for a single crane with crane
trolley positioned to produce the maximum
load effect for the element in consideration;
crane dead load includes weight of the crane
bridge, end trucks, and trolley
Cdm =
crane dead load for multiple cranes with
crane bridges and crane trolleys positioned
to produce the maximum load effect for the
element in consideration
Ci =
Crane vertical impact forces due to a single
crane in one aisle only
Cls =
longitudinal crane tractive loads from a sin-
gle crane
Css = crane side thrust from a single crane
Cvm =
crane lifted load for multiple cranes with the
crane bridges and crane trolleys positioned to
produce the maximum load effect for the ele-
ment in consideration
Cvs =
crane lifted load for a single crane with crane
trolley positioned to produce the maximum
load effect for the element in consideration
Cvs(min) =
crane lifted load for a single crane in one isle
only with crane trolley positioned to produce
the minimum load effect for the element in
consideration

Chapter 13
Structural Systems in Crane Buildings
13.1 ROOF SYSTEMS
The inclusion of cranes in an industrial building will typi-
cally not affect the basic roof covering system. Crane build-
ings will “move,” and any aspect in the roof system that
might be affected by such a movement must be carefully
evaluated. This means close examination of details (e.g.,
flashings, joints, etc.).
A difference in the roof support system design for
crane buildings, as opposed to industrial buildings without
cranes, is that a roof diaphragm system should only be used
after careful consideration of localized forces that may be
imparted into the diaphragm from the crane forces. Whereas
wind loads apply rather uniformly distributed forces to the
diaphragm, cranes forces are localized and cause concen-
trated repetitive forces to be transferred from the frame to
the diaphragm. These concentrated loads combined with the
cyclical nature of the crane loads (fatigue) should be care-
fully examined before selecting a roof diaphragm solution.
13.2 WALL SYSTEMS
The special consideration that must be given to wall systems
of crane buildings relates to movement and vibration. Col-
umns are commonly tied to the wall system to provide brac-
ing to the column or to have the column support the wall.
(The latter is applicable only to lightly loaded columns.) For
masonry and concrete wall systems, it is essential that proper
detailing be used to tie the column to the wall. Figure 13-1
illustrates a detail that has worked well for masonry walls.
The bent anchor rod has flexibility to permit movement
perpendicular to the wall but is “stiff” parallel to the wall,
enabling the wall to brace the column about its weak axis.
If a rigid connection is made between the column and the
wall and crane movements and vibrations are not accounted
for, wall distress is inevitable. The use of the wall as a lat-
eral bracing system for columns should be avoided if future
expansion is anticipated.
13.3 FRAMING SYSTEMS
The same general comments given previously for indus-
trial buildings without cranes apply to crane buildings as
well. However, the most economical framing schemes are
normally dictated by the crane. Optimum bays are usually
smaller for crane buildings than buildings without cranes
and typically fall into the 25- to 30-ft range. This bay size
permits the use of rolled shapes as crane runways for lower
load crane sizes. Larger main bay sizes of 50 to 60 ft that
utilize wind columns is more economical when deep founda-
tions and heavy cranes are specified.
The design of framing in crane buildings must include
certain serviceability considerations that are used to con-
trol relative and absolute lateral movements of the runways
by controlling the frame and bracing stiffness. The source
producing lateral movement is either an external lateral
load (wind or earthquake) or the lateral load induced by the
operation of the crane. The criteria are different for pendant-
operated versus cab-operated cranes because the operator
rides with the crane in the latter case. In crane bays with
gabled roofs, vertical roof load can cause spreading of the
eaves and thus spreading of the crane runways. Conversely,
eccentrically bracketed runways supported on building col-
umns can result in inward tilting of the columns due to the
crane loading. This would cause an inward movement of the
runways toward each other. Lastly, the crane tractive force
can cause longitudinal movement of the runway either by
Fig. 13-1. Masonry wall anchorage.

torsion in the supporting columns where brackets are used
or flexing of the frame if rigid frame bents are used for the
runway columns. Longitudinal runway movement is rarely a
problem where braced frames are used.
Recommended serviceability limits for frames supporting
cranes (West et al., 2003):
1. Pendant-operated cranes: The frame drift should be
less than the runway height over 100 based on 10-year
winds or the crane lateral loads on the bare frame.
While this limit has produced satisfactory behavior,
the range of movements should be presented to the
building owner for review because they may be per-
ceived as too large in the completed building.
2. Cab-operated cranes: The frame drift should be less
than the runway height over 240 and less than 2 in.
based on 10-year wind or the crane lateral loads on the
bare frame.
3. All top-running cranes: The relative inward move-
ment of runways toward each other should be less
than a 2-in. shortening of the runway-to-runway
dimension. This displacement would be due to crane
vertical static load. Relative outward movement of
runways away from each other should not be more
than an increase of 1 in. in the dimension between
crane runways. The loading inducing this displace-
ment would vary depending on the building location.
In areas of roof snow load less than 12 psf, no snow
load need be taken for this serviceability check. In
areas of roof snow load between 12 psf and 30 psf,
it is suggested that 50% of the roof snow load should
be used. Lastly, in areas where the snow load exceeds
30 psf, 75% of the roof snow loads should be used.
The discussion of serviceability limits is also presented in
more detail in AISC Design Guide 3.
In addition to the preceding serviceability criteria, it is
recommended that office areas associated with crane build-
ings should be isolated from the crane building so vibration
and noise from the cranes is minimized in the office areas.
13.4 BRACING SYSTEMS
13.4.1 Roof Bracing
Roof bracing is very important in the design of crane build-
ings. The roof bracing allows the lateral crane forces to be
shared by adjacent bents. This sharing of lateral load reduces
the column moments in the loaded bents. This is true for
all framing schemes (i.e., rigid frames of shapes, plates,
and trusses, or braced frames). It should be noted, however,
that in the case of rigid-frame structures, the moments in
the frame cannot be reduced to less than the wind-induced
moments.
Figures 13-2, 13-3, and 13-4 illustrate the concept of
using roof bracing to induce sharing of lateral crane loads in
the columns. For wind loading, all frames and columns are
displaced uniformly as shown in Figure 13-2. For a crane
building without roof bracing, the lateral crane loads are
transmitted to one frame line (Figure 13-3), causing signifi-
cant differential displacement between frames. The addition
of roof bracing will create load sharing. Columns adjacent
to the loaded frame will share in the load, thus reducing dif-
ferential and overall displacement (Figure 13-4). Angles or
tees will normally provide the required stiffness for this sys-
tem. Additional information on load sharing is available in
Chapter 16.
Fig. 13-2. Uniform displacement due to wind. Fig. 13-3. Displacement of unbraced Fig. 13-4. Displacement of braced
frames due to crane lateral loads. frames due to crane lateral loads.

13.4.2 Wall Bracing
It is important to trace the longitudinal crane forces through
the structure in order to ensure proper wall and crane bracing
(wall bracing for wind and crane bracing may, in fact, be the
same braces).
For lightly loaded cranes, wind bracing in the plane of the
wall may be adequate for resisting longitudinal crane forces,
as shown in Figure 13-5. For large longitudinal forces, the
bracing will likely be required to be located in the plane of
the crane rails, as shown in Figure 13-6.
For the bracing arrangement shown in Figure 13-5, the
crane longitudinal force line is eccentric to the plane of the
X-bracing. The crane column may tend to twist if the hori-
zontal truss is not provided. Such twisting will induce addi-
tional stresses in the column. The designer should calculate
the stresses due to the effects of the twisting and add these
stresses to the column axial and flexural stresses. A torsional
analysis can be performed to determine the stresses caused
by twist, or as a conservative approximation, the stresses can
be determined by assuming that the twist is resolved into a
force couple in the column flanges as shown in Figure 13-7.
The bending stresses in the flanges can be calculated from
the flange forces. In order to transfer the twist force, Pe, into
the two flanges, stiffeners may be required at the location of
the force, P.
Fig. 13-5. Wall bracing for cranes. Fig. 13-6. Vertical bracing for heavy cranes.
The following criteria will typically define the longitudi-
nal crane force transfer:
1. For small longitudinal loads (up to 4 kips), use of
wind bracing is typically efficient where columns are
designed for the induced eccentric load.
2. For medium longitudinal loads (4 kips to 8 kips), a
horizontal truss is typically required to transfer the
force to the plane of X-bracing.
3. For large longitudinal loads (more than 8 kips), brac-
ing in the plane of the longitudinal force is typically
the most effective method of bracing. Separate wind
X-bracing on braced frames may be required due to
eccentricities.
Normally the X-bracing schemes resisting these horizon-
tal crane forces are best provided by angles or tees rather
than rods. In cases where aisles must remain open, portal
type bracing may be required in lieu of designing the column
for weak axis bending, as shown in Figure 13-8.
It should be noted that portal bracing will necessitate a
special design for the horizontal (girder) member, and the
diagonals will take a large percentage of the vertical crane
forces. This system should only be used for lightly loaded,
low-fatigue situations. The system shown in Figure 13-9
could be used as an alternate to Figure 13-8. Additional
details on connections and bracing can be found in the AISC
Manual.

Fig. 13-7. Eccentric column forces.
Fig. 13-8. Portal crane runway bracing.
Fig. 13-9. Modified portal crane runway bracing.

Chapter 14
Crane Runway Design
Strength considerations for crane girder design are primarily
controlled by fatigue for CMAA 70 Class E and F cranes
and, to some degree, for Class D cranes. Wheel loads, wheel
spacing, and girder span are required for the design of crane
girders. The expense of crane girder construction typically
increases when built-up shapes are required. Fatigue restric-
tions are more severe for built-up shapes. The difference
between a rolled shape versus a built-up member using con-
tinuous fillet welds is a reduction in the allowable fatigue
stress.
The following summary of crane girder selection criteria
may prove helpful.
1. Light cranes and short spans: Use a wide-flange
beam.
2. Medium cranes and moderate spans: Use a wide-
flange beam, and if required, reinforce the top flange
with a channel or angles.
3. Heavy cranes and longer spans: Use a plate girder
with a horizontal truss or solid plate at the top flange.
4. Limit deflections under crane loads as follows (West
et al., 2003):
Vertical deflection of the crane beam due to wheel
loads (no impact)
• L/600 for light and medium cranes: CMAA 70
Classes A, B, and C
• L/800 for light and medium cranes: CMAA 70
Class D
• L/1000 for mill cranes: CMAA 70 Classes E and F
Lateral deflection of the crane beam due to crane lat-
eral loads
• L/400 for all cranes
14.1 CRANE RUNWAY BEAM DESIGN
PROCEDURE
As previously explained, crane runway beams are subjected
to both vertical and horizontal forces from the supported
crane system. Consequently, crane runway beams must be
designed for combined bending about both the x and y axis.
Salmon et al. (2008) and Gaylord et al. (1991) point out
that the equations presented in the AISC Specification for
lateral-torsional buckling strength are based upon the load
being applied at the elevation of the neutral axis of the beam.
If the load is applied above the neutral axis (for instance,
at the top flange of the beam as is the case with crane run-
way beams), lateral-torsional buckling resistance is reduced.
The lateral loads from the crane system applied at the top
flange level also generate a twisting moment on the beam.
When vertical and lateral loads are applied simultaneously,
these two effects are cumulative. To compensate for this, it
is common practice to assume the lateral loads due to the
twisting moment are resisted by only the top flange. With
this assumption, Salmon et al. and Gaylord et al. both sug-
gest that the lateral stability of a beam of this type subject
to biaxial bending is otherwise typically not affected by
the weak-axis bending moment, My. Examples provided by
these references are for relatively short beam lengths. In the
earlier editions of this Design Guide, the procedure recom-
mended by Salmon et al. and Gaylord et al. was used for
combined loading; however, the author is not aware of any
significant research on this procedure for runway girders
with varieties of shapes and spans and thus recommends that
theAISC Specification axial load and bending moment inter-
action equations be used.
Another criterion related to crane runway beam design
referred to in AISC Specification Section J10.4 is web side-
sway buckling. This criterion is included to prevent buck-
ling in the tension flange of a beam where flanges are not
restrained by bracing or stiffeners and are subject to concen-
trated loads. This failure mode may be predominant when
the compression flange is braced at closer intervals than the
tension flange or when a monosymmetric section is used
with the compression flange larger than the tension flange
(e.g., wide-flange beam with a cap channel). A maximum
concentrated load is used as the limiting criterion for this
buckling mode.
This criterion does not currently address beams subjected
to simultaneously applied multiple wheel loads. The author
is not aware of any reported problems with runway beams
that are designed using these criteria with a single wheel
load.
For crane runway beams, the following design procedure
is recommended as both safe and reasonable where fatigue
is not a factor. See Example 14.1.1 for ASD and Exam-
ple 14.1.2 for LRFD.
1. Compute the required moments of inertia, Ix and Iy, to
satisfy deflection criteria: L/600 to L/1,000 for verti-
cal deflection and L/400 for lateral deflection.

2. Position the crane to produce the worst loading condi-
tion. This can be accomplished using the equations
found in AISC Manual Part 3 for cranes with two-
wheel end trucks on simple spans. For other wheel
arrangements, the maximum moment can be obtained
by locating the wheels so that the center of the span
is midway between the resultant of the loads and the
nearest wheel to the resultant. The maximum moment
will occur at the wheel nearest to the centerline of the
span. For continuous spans, the maximum moment
determination is a trial and error procedure. Use of a
computer for this process is recommended.
3. Calculate the required bending moments, Mrx and
Mry, including the effects of impact. Many engineers
determine Mry by applying the lateral crane forces
to the top flange of the runway beam. AIST TR-13
requires that the lateral force be increased because the
force is applied to the top of the rail. This eccentricity
of lateral load increases the magnitude of the lateral
force to the top flange and requires consideration of
a corresponding bottom flange lateral force and bend-
ing moment in the opposite direction.
4. For sections without cap channels, select a trial sec-
tion ignoring lateral load.
5. To account for the weak-axis effects, select a section
with wide flanges and several sizes larger than pro-
vided by Mrx alone. For sections with cap channels, the
Appendix tables may be of assistance. If ASTM A36
cap channels are used on ASTM A992 steel beams,
then lateral-torsional buckling requirements must be
based on the ASTM A36 material, as well as the weak
axis strength.
6. Traditionally, Equations 14-1a and 14-1b have been
used to determine the strength of runway girders. The
author recommends that these be used when the AIST
TR-13 requirements are not specified.
M
M
M
M
1.0
rx
nx b
ry
ny b
Ω
+
Ω
≤ (ASD) (14-1a)
M
M
M
M
1.0
rx
b ny
nx
ry
b
ϕ
+
ϕ
≤ (LRFD) (14-1b)
where
Mnx, Mny =
nominal moments about the x- and
y-axis, respectively, kip-in.
Mny =
nominal moment based on the top-
flange area of the section about the
y-axis. For sections with cap channels,
Mny is the nominal moment based on
the channel and top-flange area, kip-in.
ϕb = 0.90
Ωb = 1.67
When designing the runway girder for bending about
the x-axis only, include the impact in the calculation
for Mrx.
For assistance in calculations, listed in Appendix
Table A-1 are the following:
It =
moment of inertia of the top flange and cap
channel about the y-axis of the combined sec-
tion, in.4
Ix =
moment of inertia about the x-axis of the
combined section, in.4
Syt =
elastic section modulus of the top flange and
cap channel about the y-axis of the combined
section, in.3
S1 =
elastic section modulus referred to the ten-
sion flange of the combined section, in.3
S2 =
elastic section modulus referred to the com-
pression flange of the combined section, in.3
Zyt =
plastic section modulus of the top flange and
cap channel about the y-axis of the combined
section, in.3
Zx =
plastic section modulus about the x-axis, in.3
y1 =
distance from the bottom flange to the section
centroid, in.
7. Check the section with respect to web sidesway buck-
ling as described in AISC Specification Section J10.4.
8. When cap channels are used, design the weld connect-
ing the channel to the flange.
In this procedure, the checks do not incorporate the force
in the runway beam due to the longitudinal tractive force.
ASCE/SEI 7-16 does not provide load combinations spe-
cifically for the runway force combinations; however, as
noted in the preceding text, AIST TR-13 does include two
load cases that incorporate the longitudinal tractive force.
Typically, the longitudinal force in the runway beam can be
checked based on the full cross-sectional area of the beam.
In the majority of cases, the required force divided by the
available strength level is so low that it can be neglected.
In selecting a trial rolled-shape section, it may be helpful
to recognize that the ratios shown in Table 14-1 exist for
various W-shapes without cap channels.
For assistance in calculating the lateral-torsional buckling
strength for sections with cap channels, listed in Appendix
Table A-2 are the following:
FL =
magnitude of flexural stress in compression flange
at which flange local buckling or lateral-torsional
buckling is influenced by yielding, ksi

Lp =
limiting unbraced length for the limit state of yield-
ing, in.
Lr =
limiting unbraced length for the limit state of inelas-
tic lateral-torsional buckling, in.
ho = distance between the flange centroids, in.
rt =
radius of gyration of the top flange and channel
about the y-axis of the combined section, in.
When fatigue is a consideration, the procedure should be
altered so the live load stress range for the critical case does
not exceed the fatigue allowable stress range determined in
accordance with AISC Specification Appendix 3.
Table 14-1.
Z
Z
x
y
Ratios for W-Shapes
W-Shape
Z
Z
x
y
W8 through W16 3 to 8
Example 14.1.1—Crane Runway Girder Design Example (ASD)
Given:
Use ASCE/SEI 7-16 to design an ASTM A992 W-shape runway girder for a CMAA Class B crane. Assume no reduction in
allowable stress due to fatigue. The design parameters are:
Type of control: cab-operated
Crane capacity = 20 tons (40 kips)
Bridge weight = 57.2 kips
Combined trolley and hoist weight = 10.6 kips
Maximum wheel load (without impact) = 38.1 kips
ASCE 100 lb/yard rail = 34 plf
Combined clamps and electrification weight = 16 plf
Bridge span = 70 ft
Runway girder span = 30 ft
Wheel spacing = 12 ft
Solution:
From AISC Manual Table 2-4, the material properties for the girder are:
ASTM A992
Fy = 50 ksi
Fu = 65 ksi
The critical wheel location with regards to bending moment is shown in Figure 14-1. The wheel location to determine approxi-
mate deflection is shown in Figure 14-2.
The required flexural strength of the runway girder is determined using AISC Manual Table 3-23, Case 44.
a = 12 ft
l = 30 ft
l
0.586 0.586 30 ft
17.6 ft
( )
=
=

Because a = 12 ft 17.6 ft:
M
P
l
l
a
P
P
2 2
2 30 ft
30 ft
12 ft
2
9.60 kip-ft
max
2
2
( )
= −
⎛
⎝
⎞
⎠
= −
⎛
⎝
⎞
⎠
=
Calculate the deflection assuming the wheel locations shown in Figure 14-2. Note that this wheel location will slightly underesti-
mate the deflection as compared to a solution based on the wheel locations shown in Figure 14-1. From AISC Manual Table 3-23,
Case 9, with a = 9 ft:
Pa
EI
l a
P
I
P
I
24
3 4
(9 ft)
24 29,000 ksi
3 30 ft 4 9 ft 1,728 in. /ft
53.1
in.
max
2 2
2 2 3 3
( )
( )
( ) )
(
( ) ( )
= −
Δ
= −
⎡
⎣
⎤
⎦
=
Vertical Deflection
Using the vertical deflection criterion of L/600, the allowable vertical deflection and required x-axis moment of inertia are:
30 ft 12 in./ft
600
0.600 in.
allow
( )( )
=
Δ
=
Fig. 14-1. Critical wheel location for bending.
Fig. 14-2. Critical wheel location for approximate deflection.

I
P
53.1
53.1 38.1 kips
0.600 in.
3,370 in.
x req d
allow
'
4
( )
=
Δ
=
=
AISC Manual Table 1-1 is used to select the crane runway girder. Try a W24×131.
=
I o.k.
4,020 in. 3,370 in.
x
4 4
Horizontal Deflection
From ASCE/SEI 7, Section 4.9.4, the lateral force on a crane runway beam with an electrically powered trolley is 20% of the
sum of the rated capacity of the crane and the weight of the hoist and trolley. The total lateral load can be distributed equally to
the four wheels on the crane. The maximum horizontal load per wheel is:
0.20
40 kips 10.6 kips
4 wheels
2.53 kips
+
⎛
⎝
⎞
⎠
=
Using horizontal deflection criterion of L/400, the allowable horizontal deflection is:
30 ft 12 in./ft
400
0.900 in.
allow
( )( )
=
Δ
=
The required y-axis moment of inertia for the top flange, Iy-top flg, is:
I
P
53.1
53.1 2.53 kips
0.900 in.
149 in.
y top flg
allow
-
4
( )
=
Δ
=
=
And the required y-axis moment of inertia for the entire section is:
I I
2
2 149 in.
298 in.
y req d y top flg
- ' -
4
4
( )
=
=
=
From AISC Manual Table 1-1, for a W24×131:
I 340 in. 298 in.
y
4 4
= o.k.
Flexural Strength
Calculate the required x- and y-axis flexural strength, Mrx and Mry.
The combined weight of the rail, clamps, and electrification is:
34 plf 16 plf 1 kip /1,000 lb 0.050 kip/ft
( )( )
+ =

The girder self-weight is:
131 plf 1 kip/1,000 lb 0.131 kip/ft
( )( ) =
For x-axis bending acting alone, the impact factor of 1.25 is applied to the crane wheel load, and the required x-axis flexural
strength is:
M 9.60 38.1 kips 1.25
0.050 kip/ft 0.131 kip/ft 30 ft
8
478 kip-ft
rx
2
( )
( )
( )
( )
= +
+
=
For x-axis bending acting in combination with lateral thrust (no impact):
M 9.60 38.1 kips
0.050 kip/ft 0.131 kip/ft 30 ft
8
386 kip-ft
rx
2
( )
( )( )
= +
+
=
The required y-axis flexural strength is:
M 9.60 2.53 kips
24.3 kip-ft
ry ( )
=
=
From AISC Manual Table 6-2, the available x-axis flexural strength for a W24×131 with Lc = 30 ft is:
M
605 kip-ft
nx
b
Ω
=
Tension Flange Combined Loading
Combined loading on the tension flange is checked using Equation 14-1a:
M
M
M
M
478 kip-ft
605 kip-ft
0
0.790 1.0
rx
nx b
ry
ny b
Ω
+
Ω
= +
= o.k.
(14-1a)
Compression Flange Combined Loading
From the AISC Manual Table 6-2, the available flexural strength about the y-axis is:
M
203 kip-ft
ny
b
Ω
=
Combined loading on the compression flange is checked using Equation 14-1a:
M
M
M
M
386 kip-ft
605 kip-ft
24.3 kip-ft
203 kip-ft
0.758 1.0
rx
nx b
ry
ny b
Ω
+
Ω
= +
= o.k.
(14-1a)

Web Sidesway Buckling
From AISC Manual Table 1-1, the geometric properties for a W24×131 are:
Sy = 53.0 in.3
bf = 12.9 in.
tf = 0.960 in.
tw = 0.605 in.
h/tw = 35.6
The clear distance between flanges less the corner radius is:
h t h t
0.605 in. 35.6
21.5 in.
w w
( )
( )( )
=
=
=
The largest unbraced length of either flange at the point of load is:
Lb = (30 ft)(12 in./ft)
= 360 in.
Because the compression flange is not restrained against rotation, AISC Specification Section J10.4(b) is used to determine the
nominal strength of the web for the limit state of web sidesway buckling:
h t
L b
35.6
360 in. 12.9 in.
1.28
w
b f ( )
=
=
Because
h t
L b
1.7,
w
b f
AISC Specification Equation J10-7 is applicable:
R
C t t
h
h t
L b
0.4 /
/
n
r w f w
b f
3
2
3
=
⎛
⎝
⎜
⎞
⎠
⎟
⎡
⎣
⎢
⎢
⎤
⎦
⎥
⎥

(Spec. Eq. J10-7)
where
Cr = 960,000 ksi when 1.5Ma My; otherwise, Cr = 480,000 ksi
Ma = required flexural strength using ASD load combinations, kip-in.
My = yield moment, kip-in.
F S
M
50 ksi 53.0 in.
2,650 kip-in.
1.5 1.5 478 kip-ft 12 in./ft
8,600 kip-in.
y y
a
3
( )
( )
( )
( )
=
=
=
=
=
Because 1.5Ma My, Cr = 480,000 ksi.
The nominal strength is:
R
480,000 ksi 0.605 in. 0.960 in.
21.5 in.
0.4 1.28
185 kips
n
3
2
3
( )( ) ( )
( )
( )
= ⎡
⎣
⎤
⎦
=

And the available strength is then:
R
1.76
185 kips
1.76
105 kips
n
=
=
=
Ω
Ω
The maximum wheel load with impact is:
38.1 kips 1.25 47.6 kips
47.6 kips 105 kips
( )( ) =
o.k.
Example 14.1.2—Crane Runway Girder Design Example (LRFD)
Given:
Repeat Example 14.1.1 for LRFD. The design parameters are found in Example 14.1.1.
Solution:
From AISC Manual Table 2-4, the material properties for the girder are:
ASTM A992
Fy = 50 ksi
Fu = 65 ksi
The LRFD load factors for crane loads are determined from ASCE/SEI 7-16.
Bridge weight: load factor = 1.2
Trolley and hoist weight: load factor = 1.2
Beam self-weight and crane rail weight: load factor = 1.2
Lifted load: load factor = 1.6
Lateral thrust: load factor = 1.6
Factored Wheel Loads
Conservatively assume that the trolley, hoist, and lifted load act over the runway beam. Because different load factors are applied
to the bridge, trolley, and hoist weights versus the lifted load, one cannot simply multiply the ASD wheel load by 1.6.
The bridge weight is distributed equally to all four wheels, and the trolley and hoist are distributed to two wheels. The factored
loads per wheel are:
P
P P
1.2
4 2
1.2
57.2 kips
4
10.6 kips
2
23.5 kips
dead
bridge trolley hoist
+
= +
⎛
⎝
⎜
⎞
⎠
⎟
= +
⎛
⎝
⎞
⎠
=
P 1.6
40 kips
2
32.0 kips
lifted load = ⎛
⎝
⎞
⎠
=
And the total factored load per wheel is:
P 23.5 kips 32.0 kips
55.5 kips
total = +
=

The factored horizontal load per wheel is:
H 1.6 0.20
10.6 kips 40 kips
4
4.05 kips
factored ( )
=
+
⎛
⎝
⎞
⎠
=
The deflection criteria are the same as calculated for Example 14.1.1.
Flexural Strength
Calculate factored Mx and My (include rail, clamps, and electrification).
The combined weight of the rail, clamps, and electrification is:
34 plf 16 plf 1 kip/1,000 lb 0.050 kip/ft
( )( )
+ =
The girder self-weight is:
131 plf 1 kip/1,000 lb 0.131 kip/ft
( )( ) =
M 9.60 55.5 kips 1.2
0.050 kip/ft 0.131 kip/ft 30 ft
8
557 kip-ft
rx
2
( )
( )( )
= +
+
=
Using an impact factor of 1.25 on the wheel loads from the lifted load, trolley, and hoist:
M 9.60 55.5 kips 1.25 1.2
0.050 plf 0.131 plf 30 ft
8
690 kip-ft
rx
2
( )
( )
( )
( )
= +
+
=
M 9.60 4.05 kips
38.9 kip-ft
ry ( )
=
=
From AISC Manual Table 6-2, for a W24×131 with Lc = 30 ft:
M 909 kip-ft
b nx
ϕ =
Tension Flange Combined Loading
Combined loading on the tension flange is checked using Equation 14-1b:
M
M
M
M
690 kip-ft
909 kip-ft
0
0.759 1.0
rx
b nx
ry
b ny
ϕ
+
ϕ
= +
= o.k.
(14-1b)
Compression Flange Combined Loading
From the AISC Manual Table 6-2, the available flexural strength about the y-axis is:
M 306 kip-ft
b ny
ϕ =

Combined loading on the compression flange is checked using Equation 14-1b:
M
M
M
M
557 kip-ft
909 kip-ft
38.9 kip-ft
306 kip-ft
0.740 1.0
rx
b nx
ry
b ny
ϕ
+
ϕ
= +
= o.k.
(14-1b)
Web sidesway buckling calculations are not repeated.
Example 14.1.3—Crane Runway Girder with Cap Channel Design Example (ASD)
Given:
Verify an ASTM A992 W30×99 crane runway girder with an ASTM A36 C15×33.9 cap channel for the design parameters pro-
vided in Example 14.1.1. The cap channel is continuously welded to the girder using 70-ksi electrodes. For comparison, use the
same dead loads from Example 14.1.1, Mrx with impact = 478 kip-ft (5,740 kip-in.), Mrx with no impact = 386 kip-ft (4,630 kip-
in.), Mry = 24.3 kip-ft (292 kip-in.)
Solution:
From AISC Manual Table 2-4, the material properties for the girder and cap channel are:
ASTM A992
Fy = 50 ksi
Fu = 65 ksi
ASTM A36
Fy = 36 ksi
Fu = 58 ksi
From AISC Manual Table 1-1, the geometric properties for the W30×99 girder are:
bf = 10.5 in.
d = 29.7 in.
kdes = 1.32 in.
tf = 0.670 in.
tw = 0.520 in.
h/tw = 51.9
From AISC Manual Table 1-5, the geometric properties for the C15×33.9 girder are:
A = 10.0 in.2
d = 15.0 in.
tw = 0.400 in.
x = 0.788 in.
From AISC Manual Table 1-19, the properties for the combined section W30×99 with C15×33.9 cap channel are:
Ix = 5,550 in.4
S1 = 300 in.3
S2 = 481 in.3
Zx = 408 in.3
y1 = 18.5 in.
From Appendix Tables A-1 and A-2, the properties of the girder with cap channel are:
FL = 31.2 ksi
It = 380 in.4
(for top flange and cap channel)
Lp = 119 in.
Lr = 457 in.

Syt = 50.6 in.3
Zyt = 69.3 in.3
ho = 29.0 in.
rt = 4.50 in.
Deflection Criteria
The deflection is checked using the limits calculated in Example 14.1.1.
Ix = 5,550 in.4
3,370 in.4
o.k.
It = 380 in.4
149 in.4
o.k.
The deflection criteria are satisfied.
Flexural Strength
AISC Specification Section F5 will be used to check flexural strength limit states in accordance with the User Note in AISC
Specification Section F4.
Compression Flange Yielding
Compression flange yielding is checked using AISC Specification Section F5.1.
M R F S
n pg y xc
= (Spec. Eq. F5-1)
Determine the bending strength reduction factor, Rpg:
R
a
a
h
t
E
F
1
1,200 300
5.7 1.0
pg
w
w
c
w y
= −
+
−
⎛
⎝
⎜
⎞
⎠
⎟ ≤

(Spec. Eq. F5-6)
a
h t
b t
w
c w
fc fc
=

(Spec. Eq. F4-12)
h d − y1 − kdes
2( )
2 29.7 in. −18.5 in. −1.32 in.
19.8 in.
c
( )
=
=
=
bfc = bf = 15.0 in. (using the channel depth as bf)
tfc = tf = 0.670 in.
a
19.8 in. 0.520 in.
15.0 in. 0.670 in.
1.02
w
( )( )
( )( )
=
=
With the variables known, Rpg can be calculated as:
R 1
1.02
1,200 300 1.02
19.8 in.
0.520 in.
5.7
29,000 ksi
50 ksi
1.07 1.0
1.0
pg
( )
= −
+
−
⎛
⎝
⎜
⎞
⎠
⎟ =
=
The nominal flexural strength for the limit state of compression flange yielding is:
M 1.0 50 ksi 481 in.
24,100 kip-in.
n
3
( )
( )
=
=

And the available flexural strength for the limit state of compression flange yielding is:
M
M
24,100 kip-in.
1.67
Ω
14,400 kip-in. 5,740 kip-in.
n
rx
=
= = o.k.
Lateral-Torsional Buckling
AISC Specification Section F5.2 is used to determine the nominal flexural strength for the limit state of lateral-torsional buckling.
Mn = RpgFcrSxc (Spec. Eq. F5-2)
Lb = (30 ft)(12 in./ft)
= 360 in.
Lp = 119 in.
Lr = 457 in.
Because Lp Lb Lr, AISC Specification Equation F5-3 is applicable:
F F F
L L
L L
F
= C 0.3
cr b y y
b p
r p
y
( )
−
−
−
⎛
⎝
⎜
⎞
⎠
⎟
⎡
⎣
⎢
⎢
⎤
⎦
⎥
⎥
≤

(Spec. Eq. F5-3)
= 1.0 50 ksi 0.3 50 ksi
360 in. 119 in.
457 in. 119 in.
39.3 ksi 50 ksi
( )( )
−
−
−
⎛
⎝
⎞
⎠
⎡
⎣
⎢
⎤
⎦
⎥
=
The nominal flexural strength for the limit state of lateral-torsional buckling is:
M 1.0 39.3 ksi 481 in.
18,900 kip-in.
n
3
( )
( )
=
=
And the available flexural strength for the limit state of lateral-torsional buckling is:
M
M
18,900 kip-in.
1.67
11,300 kip-in. 5,740 kip-in.
n
rx
Ω
=
= = o.k.
Flexural Strength of the Top Flange with Lateral Loads
The top flange is compact and thus Mp = FyZyt, where Zyt is the plastic modulus of the cap channel and the top flange.
M 50 ksi 69.3 in.
3,470 kip-in.
p
3
( )
( )
=
=
M 3,470 kip-in.
1.67
2,100 kip-in.
p
Ω
=
=

Check Top Flange for Biaxial Bending
M
M
M
M
4,630 kip-in.
11,300 kip-in.
292 kip-in.
2,100 kip-in.
0.55 1.0
rx
nx b
ry
ny b
Ω
+
Ω
= +
= o.k.
(14-1a)
Tension Flange Yielding (including impact)
Because S1 S2, AISC Specification Equation F5-10 is applicable:
M F S
n y 1
=
50 ksi 300 in.
15,000 kip-in.
3
( )
( )
=
=
(Spec. Eq. F5-10)
M
M
15,000 kip-in.
1.67
8,980 kip-in. 5,740 kip-in.
n
rx
Ω
=
= = o.k.
Web Sidesway Buckling
The clear distance between flanges less the fillet corner radius is:
h = tw(h/tw)
= (0.520 in.)(51.9)
= 27.0 in.
The largest unbraced length of either flange at the point of load is:
Lb = 360 in.
Because the compression flange is not restrained against rotation, AISC Specification Section J10.4(b) is applicable.
h t
L b
51.9
360 in. 10.5 in.
1.51
w
b f
( )
( ) ( )
=
=
Because (h/tw)/(Lb/bf) 1.7, AISC Specification Equation J10-7 is used to determine the nominal strength:
R
C t t
h
h t
L b
0.4 /
/
n
r w f w
b f
3
2
3
=
⎛
⎝
⎜
⎞
⎠
⎟
⎡
⎣
⎢
⎢
⎤
⎦
⎥
⎥

(Spec. Eq. J10-7)
M F S
M
50 ksi 50.6 in.
2,530 kip-in.
1.5 1.5 478 kip-ft 12 in./ft
8,600 kip-in.
y y yt
a
3
( )
( )
( )
( )
=
=
=
=
=
Because 1.5Ma My, Cr = 480,000 ksi.

The nominal strength is:
R
480,000 ksi 0.520 in. 0.670 in.
27.1 in.
0.4 1.52
86.5 kips
n
3
2
3
( )( ) ( )
( )
( )
= ⎡
⎣
⎤
⎦
=
And the available strength is then:
R 86.5 kips
1.76
Ω
49.1 kips
n
=
=
38.1 kips 1.25 47.6 kips
47.6 kips 49.1 kips
( )( ) =
o.k.
A W30×99 with a C15×33.9 cap channel is adequate.
Weld Requirements between the Cap Channel and the Top Flange
As indicated in Section 11.2, continuous welds should be used to connect the cap channel to the beam top flange. The cap channel
always extends the full length of the beam and thus is not a partial-length cover plate; however, it is appropriate to develop the
cap channel according to the requirements of AISC Specification Section F13.3, “For welded cover plates, the welds connecting
the cover plate termination to the beam or girder shall have continuous welds along both edges of the cover plate in the length a′,
defined in the following, and shall develop the cover plate’s portion of the available strength of the beam or girder at the distance
a′ from the end of the cover plate.” For this case, the cap channel must be developed as required by the location of the required
moment. This requirement is typically satisfied by using the strength of materials equation, q = VQ/I, where
I = moment of inertia of the composite section, in.4
Q =
area of the cap channel multiplied by the distance from the composite neutral axis to the weak axis centroid of the chan-
nel, in.3
V = shear from the vertical crane loads at the end of the runway beam, kips
q = required weld shear, kip/in.
The maximum shear at the end of the beam:
Wheel loads 47.6 kips
18 ft 47.6 kips
30 ft
76.2 kips
( )
( )
= +
=
Dead load
0.050 kip/ft 0.099 kip/ft 0.0339 kip/ft 30 ft
2
2.74 kips
( )( )
=
+ +
=
V =75.0 kips + 2.74 kips
= 77.7 kips
Q A y
ch
=
y d x y
total 1
= − −

The total depth of the section, dtotal, equals the depth of the W30 plus the web thickness of the C15:
d 29.7 in. 0.400 in.
= 30.1 in.
total = +
y 30.1 in. 0.788 in. 18.5 in.
10.8 in.
= − −
=
Q 10.0 in. 10.8 in.
108 in.
2
3
( )( )
=
=
q
77.7 kips 108 in.
5,550 in.
1.51 kip/in.
0.756 kip/in./side
3
4
( )
( )
=
=
=
The required weld size is determined using AISC Manual Equation 8-2b:
R
Dl
0.928 kip/in.
n
( )
Ω
=

(Manual Eq. 8-2b)
D
q
0.928 kip/in.
0.756 kip/in.
0.928 kip/in.
0.815 sixteenths
=
=
=
The W30×99 has a flange thickness of 0.670 in., and the C15×33.9 has a web thickness of 0.400 in.
From AISC Specification Table J2.4, the minimum size of the fillet weld is x in.
Use a continuous x-in. fillet weld.
Because the fillet welds will develop the channel for the vertical loads, they will also satisfy the weld requirements for the smaller
lateral loads.
Example 14.1.4—Crane Runway Girder with Cap Channel Design Example (LRFD)
Given:
Repeat Example 14.1.3 using LRFD. The design parameters are found in Example 14.1.1. From Example 14.1.2, the factored
moments are Mrx with impact = 690 kip-ft (8,280 kip-in.), Mrx with no impact = 557 kip-ft (6,680 kip-in.), and Mry = 38.9 kip-ft
(467 kip-in.)
Solution:
The deflection criteria are the same as calculated for Example 14.1.1:
Ix = 5,550 in.4
3,370 in.4
o.k.
It = 380 in.4
149 in.4
o.k.
Flexural Strength
From the previous calculations:

Mnx = 19,000 kip-in
Mny = 3,470 kip-in.
Check Top Flange for Biaxial Bending
M
M
M
M
6,680 kip-in.
0.90 19,000 kip-in.
467 kip-in.
0.90 3,470 kip-in.
0.54 1.0 o.k.
rx
nx
ry
ny
( )
( )
( )
( )
ϕ
+
ϕ
= +
=
(14-1b)
Tension Flange Yielding (including impact)
M F S
50 ksi 300 in.
15,000 kip-in.
n y 1
3
( )
( )
=
=
=
M
M
0.9(15,000 kip-in.)
= 13,500 kip-in. 8,280 kip-in.
n
rx
ϕ =
= o.k.
Check Web Sidesway Buckling
From the previous calculations in Example 14.1.3, the nominal strength is:
R 86.5 kips
n =
The available strength is:
R
0.85
0.85 86.5 kips
73.5 kips
n ( )
ϕ =
ϕ =
=
From Example 14.1.2, the factored wheel load not including impact = 55.5 kips.
55.5 kips 1.25 69.4 kips
69.4 kips 73.5 kips
( )( ) =
o.k.
A W30×99 with a C15×33.9 cap channel is adequate, as it was in the ASD solution. The W24×131 plain beam is the most eco-
nomical solution.
Welding requirements are not repeated.

14.2 PLATE GIRDERS
Plate-girder runways can be designed in the same manner as
rolled sections, but the following items become more impor-
tant to the design.
1. Plate-girder runways are normally used in mill build-
ings where many cycles of load occur. Because they
are built-up sections, fatigue considerations are
extremely important.
2. Welding stiffeners to the girder webs may produce
a fatigue condition that would require a reduction
in stress range (Reemsynder, 1978). Thickening the
girder web so that stiffeners are not required (except
for the bearing stiffeners that are located at points of
low flexural stress) may provide a more economical
solution. However, in recent years, numerous cases
of fatigue cracks at the junction of the top flange of
the girder and the web have been noted. These cracks
have been due to:
a. The rotation of the top flange when the crane rail
was not directly centered over the web as shown in
Figure 14-3.
b. The presence of residual stresses from the welding
of the flange and stiffeners to the web.
c. Localized stresses under the concentrated wheel
loads.
The presence or absence of stiffeners affects items 2a
and 2c. If intermediate stiffeners are eliminated or reduced,
the problem of eccentric crane rail location becomes more
severe. If intermediate stiffeners are provided, CJP welds
should be used to connect the top of the stiffener to the
underside of the top flange. At the tension flange, the stiffen-
ers should be terminated not less than four times or more
than six times the web thickness from the toe of the web-to-
flange weld.
Fig. 14-3. Rail misalignment.
Shown in Figures 14-4 through 14-7 are details that per-
tain to heavy crane runway installations. The solution for
different depth girders as shown in Figure 14-6 can be prob-
lematic for potential girder replacement. An alternative is
to cope the deeper girder so that both girders are supported
directly on the column top. The tension rod shown in Fig-
ure 14-7 provides an additional load path (other than the
bolts in combined tension and shear) for the stop forces and
may be a good detail to use with high-speed cranes.
The difference in weld and stiffener detailing between
olderAISC publications and the stiffeners shown here are the
result of revised detailing techniques for fatigue conditions.
14.3 SIMPLE SPAN VERSUS CONTINUOUS
RUNWAYS
The decision to use simple-span or continuous runway crane
girders has been debated for years. In general, continuous
girders should not be used unless absolutely necessary.
Following is a brief list of advantages of each system. It is
clear that each can have an application.
1. Advantages of simple-span girders:
a. Much easier to design for various load
combinations.
b. Typically unaffected by differential settlement of
the supports.
c. More easily replaced if damaged.
d. More easily reinforced if the crane capacity is
increased.
2. Advantages of continuous girders:
a. Continuity reduces deflections that quite often
control.
b. Result in lighter-weight shapes and a savings in
steel cost when fatigue considerations are not a
determining factor.

Fig. 14-4. Girder splice.
Fig. 14-5. Girder tieback.

Fig. 14-6. Section at different-depth crane girders.
Fig. 14-7. Heavy-duty crane stop.

Continuous girders should not be used if differential set-
tlement of the supports is of the magnitude that could cause
damage to the continuous members. Also, when continuous
girders are subjected to fatigue loading and have welded
attachments on the top flange (rail clips), the allowable stress
range is reduced considerably. Any advantage may therefore
be eliminated.
A comparison of several runway beam designs is shown
in Figure 14-8. For spans varying from 20 to 30 ft, 50-ksi
beams were designed for a 4-wheel, 10-ton crane, with a
70-ft bridge for continuous (two-span) versus simple-span
conditions. In these examples, deflection did not control.
Fatigue was not considered. The curves represent the trends
for heavier cranes as well. In general, two-span continuous
crane girders could save about 18% in weight over simply
supported girders.
14.4 LATERAL LOAD-RESISTING MEANS
There are several ways to resist the lateral crane loads for
the design of runway girders. The three main methods are
briefly discussed in the following sections.
14.4.1 Cap Channels, Cap Plates, or Angles Welded to
the Top Flange
Cap channels are often used to control lateral deflections and
to control the stresses due to lateral loads. For light-duty and
lightweight cranes (less than 5 ton), cap channels may not be
required. Studies have found that a steel savings of approxi-
mately 30 lb/ft is required to justify the cost of welding a
cap to a structural shape, and thus, their use is many times
not justified. As discussed in Section 11.2, “Crane Runway
Fatigue Considerations,” for CMAA Class E and F, cranes
cap channels and cap plates are not recommended.
Fig. 14-8. Runway beam design comparison.
14.4.2 Oversized Top Flange
When plate girders are designed, the top flange can be
designed to provide the necessary strength for vertical bend-
ing and lateral bending.
14.4.3 Backup Trusses and Apron Plates
For runway girders with spans in the range of 60 ft or more,
the best solution is to design a horizontal truss or a horizon-
tal apron plate to resist the lateral crane loads. The truss or
apron plate is supported vertically by the runway girder on
one edge and a vertical truss on the other edge.
14.5 RUNWAY BRACING CONCEPTS
Mueller (1965) wrote an excellent paper on the subject of
crane-girder bracing. As illustrated in Figure 2 in the Muel-
ler paper (repeated here as Figure 14-9), improper detailing
at the end-bearing condition could lead to a web tear in the
end of the crane girder. Although Figure 14-9 indicates rivets
in the plate, the same situation could exist if bolts were used.
The detail shown in Figure 14-10 has been used to elimi-
nate this problem for light-crane systems. The welded detail
should not be used other than for CMAA crane Classes A
and B. The tiebacks as shown restrict girder end rotation and
thus can crack due to fatigue. The details previously shown
in Figures 14-5 and 14-6 represent similar details for heavy
cranes. Use of this detail allows end rotation and yet prop-
erly transfers the required lateral forces into the column.
A common method of bracing the crane girder is to pro-
vide a horizontal truss (lacing) or a horizontal plate to con-
nect the crane girder top flange to an adjacent structural
member as previously illustrated in Figures 13-5 and 13-6.
An advantage of using the horizontal plate is that it can be
used as a walkway for maintenance purposes.

Fig. 14-9. Improper girder connection detail.
Fig. 14-10. Proper tieback detail.

A critical consideration in the use of this system is using
flexible lacing in the vertical direction, enabling the crane
girder to freely deflect relative to the structural member to
which it is attached. If the lacing is not flexible, stresses will
be produced that could cause a fatigue failure of the lacing
system, thereby losing the lateral support for the girder.
AIST TR-13 requires that girders more than 36 ft in length
must have the bottom flange braced by a horizontal truss
system. The origin of this requirement is not obvious; how-
ever, it appears that compliance with the AISC Specification
web sidesway buckling equations may analytically satisfy
this requirement. The best solution to prevent web sidesway
buckling is to select a girder with a wide bottom flange.
14.6 CRANE STOPS
AIST TR-13 indicates that, “The load applied to the runway
crane stop shall be included in the design of crane runway
girders, their connections and supporting framework. The
maximum design bumper force shall be coordinated with the
crane designer and shown on the structural drawings. The
design bumper force shall be less than or equal to the maxi-
mum allowable force on the crane stop.”
Currently most crane stops are designed and supplied
using hydraulic bumpers. The AIST TR-13 Commentary
contains an example calculation for determining the forces
on the structure when hydraulic bumpers are used. An excel-
lent reference for the design criterion for hydraulic bum-
pers is “Hydraulic Bumpers for the Protection of Buildings,
Cranes and Operators from Impact Damage” (Kit, 1996).
Older bumpers were designed and supplied as spring-type
bumpers. For spring-type bumpers, the longitudinal crane
stop force acting at the center of mass of the bridge and trol-
ley may be calculated from Equation 14-3. The force on each
runway stop is the maximum bumper reaction from the iner-
tial force acting at such locations.
F
WV
gct
2
=

(14-3)
where
V =
specified crane velocity at moment of impact,
required by Specification for Electric Overhead
Traveling Cranes for Steel Mill Service, AIST TR-06
(AIST, 2018) to be 50% of full load-rated speed, ft/s
W = total weight of crane exclusive of lifted load, kips
ct =
stroke of spring at point where the crane stopping
energy is fully absorbed, ft
g = acceleration of gravity, 32.2 ft/s2
For bumper blocks of wood or rubber (commonly found
in older cranes), Equation 14-3 is not directly applicable.
Manufacturer’s literature or experience must be used for
such installations. In the absence of specific data, it is rec-
ommended that the designer assume the bumper force to be
the greater of twice the tractive force or 10% of the entire
crane weight.
14.7 CRANE RAIL ATTACHMENTS
There are four general types of anchoring devices used to
attach crane rails to crane runway beams: hook bolts, rail
clips, rail clamps, and patented clips. Details of hook bolts
and rail clamps are shown in the AISC Manual.
14.7.1 Hook Bolts
Hook bolts provide an adequate means of attachment for
light rails (40 lb–60 lb) and light-duty cranes (CMAA 70
Classes A, B, and C). The advantages of hook bolts are,
(1) they are relatively inexpensive, (2) there is no need to
provide holes in the runway beam flange, and (3) it is easy to
install and align the rail. They are not recommended for use
with heavy-duty cycle cranes (CMAA 70 Classes D, E, and
F) or with heavy cranes (greater than 20-ton lifting capac-
ity) because hook bolts are known to loosen and/or elongate.
It is generally recommended that hook bolts should not be
used in runway systems that are longer than 500 ft because
the bolts do not allow for longitudinal movement of the rail.
Because hook bolts are known to loosen in certain applica-
tions, a program of periodic inspection and tightening should
be instituted for runway systems using hook bolts. Designers
of hook bolt attachments should be aware that some manu-
facturers supply hook bolts of smaller-than-specified diam-
eter by the use of upset threads.
14.7.2 Rail Clips
Rail clips are forged or cast devices that are shaped to match
specific rail profiles. They are typically bolted to the runway
girder flange with one bolt or are sometimes welded. Rail
clips have been used satisfactorily with all classes of cranes.
However, one drawback is that when a single bolt is used,
the clip can rotate in response to rail longitudinal movement.
This clip rotation can cause a camming action thus forcing
the rail out of alignment. Because of this limitation, rail clips
should only be used in crane systems subject to infrequent
use and for runway systems less than 500 ft in length.
14.7.3 Rail Clamps
Rail clamps are a common method of attachment for heavy-
duty cycle cranes. There are two types of rail clamps—tight
and floating. Each clamp consists of an upper clamp plate
and a lower filler plate.
The lower plate is flat and roughly matches the height of
the toe of the rail flange. The upper plate covers the lower

plate and extends over the top of the lower rail flange. In the
tight clamp, the upper plate is detailed to fit tight to the lower
rail flange top, thus “clamping” it tightly in place when the
fasteners are tightened. In the past, the tight clamp had been
illustrated with the filler plates fitted tightly against the rail
flange toe. This tight fit-up was rarely achieved in practice
and is not considered to be necessary to achieve a tight-type
clamp. In the floating-type clamp, the pieces are detailed to
provide a clearance both alongside the rail flange toe and
below the upper plate. The floating type does not, in reality,
clamp the rail but merely holds the rail within the limits of
the clamp clearances. High-strength bolts are recommended
for both clamp types.
Tight clamps are typically preferred and recommended by
crane manufacturers because of concerns that the transverse
rail movement allowed in the floating type causes acceler-
ated wear on crane wheels and bearings. Floating rail clamps
may be required by crane runway and building designers to
allow for longitudinal movement of the rail, thus preventing
or at least reducing thermal forces in the rail and runway
system.
Because tight clamps prevent longitudinal rail movement,
they should not be used in runways greater than 500 ft in
length. Because floating rail clamps are frequently needed
and crane manufacturers’ concerns about transverse move-
ment are valid, a modified floating clamp is required. In such
a clamp, it is necessary to detail the lower plate to a closer
tolerance with respect to the rail flange toe. The gap between
the lower plate edge and flange toe can vary between snug
and a gap of 8 in. The 8-in. clearance allows a maximum
of 4-in. float for the system. This should not be objection-
able to crane manufacturers because this amount of float is
within normal CMAA 70 tolerances for crane spans in the
range of 50 to 100 ft—that is, spans typically encountered in
general construction. In order to provide field adjustment for
variations in the rail width, runway beam alignment, beam
sweep, and runway bolt hole location, the lower plate can
be punched with the holes off center so that the plate can be
flipped to provide the best fit. An alternative would be to use
short-slotted or oversize holes. In this case, one must rely on
bolt tightening to clamp the connection to prevent filler plate
movement.
Rail clamps are typically provided with two bolts per
clamp. Two bolts are desirable because they prevent the
camming action described in the section on rail clips. A
two-bolt design is especially recommended if clamps of the
longitudinal expansion type described previously are used.
Rails are sometimes installed with pads between the rail and
the runway beam. When this is done the lateral float of the
rail should not exceed Q in. to reduce the possibility of the
pads being worked out from under the rail.
14.7.4 Patented Rail Clips
This fourth type of anchoring device covers various patented
devices for crane rail attachment. Each manufacturer’s lit-
erature presents in detail the desirable aspects of the vari-
ous designs. In general, they are easier to install due to their
greater range of adjustment while providing the proper limi-
tations of lateral movement and allowance for longitudinal
movement. Patented rail clips should be considered as a via-
ble alternative to conventional hook bolts, clips, or clamps.
Because of their desirable characteristics, patented rail clips
can be used without restriction except as limited by the spe-
cific manufacturer’s recommendations. Installations using
patented rail clips sometimes incorporate pads beneath the
rail. When this is done, the lateral float of the rail should be
limited as in the case of rail clamps.
14.7.5 Design of Rail Attachments
The design of rail attachments is largely empirical. The
selection of the size and spacing of attachments is related to
rail size. This relation is reasonable because the rail size is
related to load.
With regard to spacing and arrangement of the attachment,
the following recommendations are given. Hooked bolts
should be installed in opposing pairs with 3 to 4 in. between
the bolts. The hooked bolt pairs should not be spaced farther
than 2 ft apart. Rail clips and clamps should be installed in
opposing pairs. They should be spaced 3 ft apart or less.
In addition to crane rail attachment, other attachments in
the form of clips, brackets, stiffeners, etc., are often attached
to the crane girder. These attachments are often added by
plant engineering personnel. Welding should only be done
after a careful engineering evaluation of its effects. Welding
(including tack welding) can significantly shorten the fatigue
life. Therefore:
1. Never weld crane rails to girders.
2. Clamp, screw, or bolt all attachments to crane girders
to avoid fatigue problems.
3. All modifications and repair work must be submit-
ted to engineering for review and approval before the
work is performed.
14.7.6 Rail Pads
One aspect of crane rail design is the use of crane rail pads.
These are generally preformed fabric pads that work best
with welded rail joints. The resilient pads will reduce fatigue,
vibration, and noise problems. Reductions in concentrated
compression stresses in the web have been achieved with the
use of these pads. Significant reductions in wear to the top of
the girder flange have also been noted. With the exception of

a few patented systems, the pads are generally not compat-
ible with floating rail installations because they can work
their way out from under the rail. Also, prior to using a pad
system, careful consideration to the cost benefits of the sys-
tem should be evaluated.
14.8 CRANE RAILS AND CRANE RAIL JOINTS
The selection of rail relates to crane considerations (basi-
cally, crane weight) and is typically made by the crane
manufacturer. Once this decision is made, the principal con-
sideration is how the rail sections are to be joined. There
are several methods to join rails, but two predominate at the
present time.
The bolted butt joint is the most commonly used rail joint.
Butt joint alignment is created with bolted splice plates.
These plates must be properly maintained (bolts kept tight).
If splice bars become loose and misaligned joints occur, a
number of potentially serious problems can result, including
chipping of the rail, bolt fatigue, damage to crane wheels,
and as a result of impact loading, increased stresses in the
girders. Girder web failures have been observed as a conse-
quence of this problem.
The welded butt joint, when properly fabricated to pro-
duce full strength, provides an excellent and potentially
maintenance-free joint. However, if repairs are necessary to
the rails, the repair procedure and consequently the down
time of plant operations is typically longer than if bolted
splices had been used. The metallurgy of the rails must be
checked to ensure use of proper welding techniques, but if
this is accomplished, the advantages can be significant. Prin-
cipal among these is the elimination of joint impact stresses
resulting in reduced crane wheel bearing wear.
Rail joints should be staggered so that the joints do not
line up on opposite sides of the runway. The amount of stag-
ger should not equal the spacing of the crane wheels, and in
no case should the stagger be less than 1 ft.
Rail misalignment is the single most critical aspect of the
development of high-impact and lateral stresses in crane
girders. Proper use and maintenance of rail attachments is
critical in this regard. Rail attachments must be completely
adjustable and yet be capable of holding the rail in align-
ment. Because the rails may become misaligned, regular
maintenance is essential to correct the problem.
14.9 RUNWAY CLEARANCES, TOP OF RAIL
ELEVATION, AND BUILDING EAVE HEIGHT
The top of crane rail (TOR) elevation and the building eave
height are established from the hook height. If the crane has
not been ordered, the crane dimensions must be approxi-
mated; otherwise, the crane data sheets can be used. A good
source for dimensional information is the Whiting Crane
Handbook (Whiting, 1967). The owner must establish the
hook height that is required. Once this is known, the TOR
can be established from the crane data sheet, and the eleva-
tion of the top of the crane can be established. OSHA Over-
head and Gantry Cranes, Subpart N, 29 of CFR 1910.179
(OSHA, 2010a), hereafter referred to as OSHA Subpart N,
requires a minimum clearance of 3 in. between the top of the
crane and the bottom of the roof members or other obstruc-
tions. Allowance must be made for piping, lights, and other
items that might be below the bottom of the roof members.
The building eave height can be established as follows:
Eave height = hook height + hook to top of crane + clear-
ance to structure + structural member depth
+ deck height
OSHA Subpart N also requires a minimum of 2 in.
between the end truck and the structural columns. When an
apron plate exists, the distance between the end truck and the
structural column should be a minimum of 18 in.

Chapter 15
Crane Runway Fabrication and Erection Tolerances
Crane runway fabrication and erection tolerances should be
addressed in the project specifications because standard tol-
erances used in steel frameworks for buildings are not tight
enough for buildings with cranes. Also, some of the required
tolerances are not addressed in standard specifications.
Tolerances for structural shapes and plates are given in the
Standard Mill Practice section of the AISC Manual. These
tolerances cover the permissible variations in geometrical
properties and are taken from ASTM specifications, AISI
steel product manuals, and producer’s catalogs. In addition
to these standards, the following should be applied to crane
runways.
The following tolerances are from AIST TR-13:
a. Sweep: Not to exceed 4 in. in a 50-ft beam length.
b. Camber: Not to vary from the camber given on the
drawings by ±4 in. in a 50-ft beam length.
c. Squareness: Within 18 in. of each girder end, the
flange is required to be free of curvature and normal
to the girder web.
Columns, base plates, and foundations should adhere to
the following tolerances:
a. Column anchor bolts should not deviate from their
theoretical location by 0.4 times the difference
between bolt diameter and hole diameter through
which the bolt passes.
b. Individual column base plates should be within ±z in.
of theoretical elevation and should be level within
±0.01 in. across the plate length or width. Paired base
plates serving as a base for double columns should be
at the same level and should not vary in height from
one to another by more than z in.
Crane runway girders and crane rails should be fabricated
and erected for the following tolerances:
a. Crane rails should be centered on the centerline of the
runway girders. The maximum eccentricity of center
of rail-to-centerline of girder should be three-quarters
of the girder web thickness.
b. Crane rails and runway girders should be installed to
maintain the following tolerances:
1. The horizontal distance between crane rails should
not exceed the theoretical dimension by ±4 in.
measured at 68°F.
2. The longitudinal horizontal misalignment from
straight of rails should not exceed ±4 in. in 50 ft
with a maximum of ±2-in. total deviation in the
length of the runway.
3. The vertical longitudinal misalignment of crane
rails from straight should not exceed ±4 in. in
50 ft measured at the column centerlines with a
maximum of ±2-in. total deviation in the length
of the runway.
Table 15-1 is taken from the MBMA Low Rise Building
Systems Manual (MBMA, 2012) and gives alternate toler-
ances. CMAA has also established crane runway installa-
tion tolerances that may be the controlling requirement for
the project. Crane suppliers may require conformance with
CMAA 70 tolerances, which differ from AIST TR-13 and
the MBMA Manual.

Table 15-1. Summary of Crane Runway Tolerances
Item Illustration Tolerance
Maximum Rate
of Change
Span A = a in. 4 in./20 ft
Straightness B = a in. 4 in./20 ft
Elevation C = a in. 4 in./20 ft
Beam-to-beam
top running
D = a in. 4 in./20 ft
Beam-to-beam
underhung
E = a in. 4 in./20 ft
Adjacent beams F = 8 in. NA

Chapter 16
Column Design
No attempt will be made to give a complete treatise on the
design of steel columns. The reader is referred to several
excellent texts on this subject: Gaylord et al. (1991) and
Salmon et al. (2008).
This section of the Guide includes a discussion of the
manner in which a crane column can be analyzed, how
the detailing and construction of the building will affect
the loads the crane column receives, and how shear and
moment will be distributed along its length. Also included
is a detailed example of a crane column design to illustrate
certain aspects of the design.
In most crane buildings the crane columns are statically
indeterminate. Typically, the column is restrained at the bot-
tom by some degree of base fixity. The degree of restraint is,
to a large extent, under the control of a designer, who may
require either a fixed base or a pinned base.
It is essential to understand that the proper design of crane
columns can only be achieved when column moments are
realistically determined. This determination requires a com-
plete frame analysis in order to obtain reliable results. Even
if a complete computer frame analysis is employed, certain
assumptions must still be made of the degree of restraint at
the bottom of a column and the distribution of lateral loads
in the structure. Further, in many cases, a preliminary design
of these crane columns must be performed either to obtain
approximate sizes for input into a computer analysis or for
preliminary cost and related feasibility studies. Simplifying
assumptions are essential to accomplish these objectives.
16.1 BASE FIXITY AND LOAD SHARING
Crane columns are constructed as bracketed, stepped, laced,
or battened, as shown in Figure 16-1. In each case, the
eccentric crane loads and lateral loads produce moments in
the column. The distribution of column moments is one prin-
cipal consideration.
For a given loading condition, the moments in a crane col-
umn are dependent on many parameters. Most parameters
(e.g., geometry, nonprismatic conditions) are readily accom-
modated in the design process using standard procedures.
However, two parameters that have a marked effect on col-
umn moments are base fixity and the amount of load sharing
with adjacent bents.
For the column configuration illustrated in Figure 16-2,
the model used for the analysis is shown in Figure 16-3. The
loading consists of a vertical crane load of 310 kips to the
left column and 100 kips to the right column. The eccentric-
ity of the vertical load to the column centroid is 1.51 ft. The
lateral crane load to each side is 23 kips. A stepped column
is used, but the same general principles apply to the other
column types. For simplicity, no roof load is used.
Base Fixity
The effect of base fixity on column moments was determined
by a first-order elastic analysis for the frame for fixed- and
pinned-base conditions. The results of the analysis shown
in Figure 16-4 demonstrate that a simple base will result in
(a) bracketed (b) stepped (c) laced (d) battened
Fig. 16-1. Column types.

Fig. 16-2. Example frame.
Fig. 16-3. Analysis model.

extremely large moments in the upper portion of the column,
and the structure will be much more flexible than a fixed-
base column.
For fixed-base columns, the largest moment is carried to
the base section of the column where it can, in the case of the
stepped column, be more easily carried by the larger section.
It is frequently argued that taking advantage of full fix-
ity cannot be achieved in any practical detail. However, the
crane-induced lateral loads on the crane column are of short
duration, and for such short-term loading, an “essentially
fixed” condition can typically be achieved through proper
design. The reduced column moments due to the fixed-
base condition provide good economy without sacrificing
stiffness.
There will be cases where subsoil conditions, existing
construction restrictions, property line limitations, etc., will
preclude the development of base fixity and the hinged base
must be used in the analysis. Although the fixed-base concept
as stated is deemed appropriate due to the short-term nature
of crane loads, for other long-duration building loads, the
assumption of full fixity may be inappropriate. The reader
is referred to an excellent article by Galambos (1960) that
deals with the effects of base fixity on the buckling strength
of frames.
Load Sharing to Adjacent Bents
If a stiff system of bracing is used (i.e., a horizontal bracing
truss as shown in Figure 16-5), then the lateral crane forces
and shears can be distributed to adjacent bents, thereby
reducing column moments. Note that such bracing does
not reduce column moments induced by wind, seismic, or
roof loads, but only the singular effects of crane loads. Fig-
ure 16-6 depicts the moment diagram in the column from a
frame analysis based on lateral crane loads being shared by
the two adjacent frames (i.e., two-thirds of the lateral sway
force is distributed to other frames). The significant reduc-
tions in moment are obvious when compared to Figure 16-4.
(Note: The “two-thirds” is an arbitrary distribution used at
this point only to illustrate the concept and the significant
advantage to be gained. The following paragraphs describe
in detail how load sharing actually occurs and how it can be
evaluated.)
Fig. 16-4. Analysis results.
Fig. 16-5. Roof plane horizontal bracing.

Consider a portion of a roof system consisting of five
frames braced as shown in Figure 16-7. The lateral crane
force will result in a reactive force at the level of the lower
chord of the roof truss, as shown in Figure 16-8. The distri-
bution of this reactive force to the adjacent frames can be
obtained by stiffness methods. This is accomplished by ana-
lyzing the horizontal bracing system as a truss on a series of
elastic supports. The supports are provided by the building
frames and have linear elastic spring constants equal to the
reciprocal of the displacement of individual frames due to
a unit lateral load, as shown in Figure 16-9. The model is
depicted in Figure 16-10. The springs are imaginary mem-
bers that provide the same deflection resistance as the frames.
This procedure has been programmed and analyzed for
many typical buildings. It is obvious that the degree of load
sharing varies and is dependent upon the relative stiffness
of the bracing to the frames; however, it was found that for
typical horizontal bracing systems, a lateral load applied to a
single interior frame will be shared almost equally by at least
five frames. This is logical because bracing of reasonable
proportions made up of axially loaded members is many
times as stiff as the moment frames that are dependent upon
the bending stiffness of their components.
A building supporting a 100-ton crane is used to illustrate
the effect of load sharing. A roof system consisting of five
frames cross-braced as shown in Figure 16-7 was analyzed
as shown in Figure 16-10 to determine the force in each
frame due to a 20-kip force applied to the center frame. The
20-kip force represents the reactive force at the elevation of
the bottom chord bracing due to the horizontal crane thrust
at the top of the crane girder. The final distribution is shown
in Figure 16-11.
Even though reasonable truss type bracing will distrib-
ute a concentrated lateral force to at least five frames, it is
recommended that load sharing be limited to three frames
(the loaded frame plus the frame to either side). The reason
Fig. 16-6. Moment diagram with load sharing. Fig. 16-7. Portion of a roof-framing plan.
Fig. 16-8. Reactive force.

Fig. 16-9. Unit lateral load.
Fig. 16-10. Computer model.
Fig. 16-11. Final force distribution.

for this conservative recommendation is that unless the hori-
zontal bracing truss members are pretensioned, they may
tend to sag even though “draw” is provided. Thus, a certain
amount of movement may occur before the truss “takes up”
and becomes fully effective in distributing the load to adja-
cent frames. The designer may conclude that if load sharing
occurs, a simple method to handle the analysis is to design
a given column for one-third the lateral load. This is wrong
and unsafe! Each individual crane column must be designed
for the full lateral force of the crane. It is only the reactive
force applied at the level of the bracing that is distributed
to the adjacent frames. The results of this analysis must be
added to or compared with the results of other analyses that
are unaffected by the load sharing—that is, gravity, wind,
and seismic loads.
The foregoing discussion was presented to explain how
load sharing works. Most engineers will determine the col-
umn moments and forces by modeling three frame lines with
the horizontal truss included in the model.
To summarize, the most economical designs will result
when the following criteria are designed into the structure:
1. Fixed-base columns.
2. Horizontal bracing truss (unless wind loads control)
such that lateral crane loads are distributed to adjacent
columns, which reduces frame drift and moments due
to drift.
3. When the roof frames are fabricated trusses, the most
economical bracing truss location is at the elevation
of the bottom chord where they are typically easier
to erect. The bottom-chord bracing system that is
required for uplift and slenderness ratio control may
also be adequate for distributing concentrated lateral
forces.
16.2 PRELIMINARY DESIGN METHODS
Preliminary design procedures for crane columns are espe-
cially helpful due to the complexity of design of these mem-
bers. Even with the widespread availability of computers,
a good preliminary design can result in substantial gains in
overall efficiency. The preceding sections have pointed out
that to obtain meaningful column moments, a frame analysis
is required. A reliable hand calculation method for prelimi-
nary design is not only helpful but essential to reduce final
design calculation time.
The frame analysis that is required to obtain an exact solu-
tion accomplishes the following:
1. It accounts for sidesway.
2. It properly handles the restraint at the top and at the
base of the column.
3. It accounts for non-prismatic member geometry.
16.2.1 Stepped Columns
What is needed for a preliminary design procedure is a
method of analysis that will provide suitable column stiff-
ness estimates so that an exact indeterminate frame analysis
procedure need be conducted only once. The model shown
in Figure 16-12(a) has been found to give good results for
crane loads, providing horizontal bracing is used in the final
design. It is a “no-sway” model, consisting of a fixed base
and supports introduced at the two points where the truss
chords intersect the column.
The moment diagram obtained from the no-sway model
for the 100-ton crane column shown in Figure 16-12(a) is
shown in Figure 16-12(b).
Comparing Figure 16-12(b) to Figure 16-6, it can be seen
that the general moment configuration is similar, and the
magnitudes of moments are almost identical for the lower
shaft. For preliminary design purposes, the two-support, no-
sway model is adequate. The two-support, no-sway model is
statically indeterminate to the second degree. Thus, even a
preliminary design requires a complex analysis and certain
other assumptions.
To obtain accurate values for moments, the effects of
the nonuniform column properties must be included in the
analysis. In doing a preliminary analysis of a stepped col-
umn, the substitution of a single top-hinge support to replace
the two supports in the two-support, no-sway mode is often
used. The single-hinged support is located at the intersection
of the bottom chord and the column.
The simplified structure is depicted in Figure 16-13.
Equations for the analysis of this member are given in
Figure 16-14.
In each case, the equation for the top shear force is given.
For the single-support assumption, the indeterminacy is
eliminated once this shear force is known. The moment
diagram for the single-hinge, no-sway column is evaluated
using the equations provided in Figure 16-14. The moment
diagram is shown in Figure 16-15.
While the variation in moment along the length is not in
good agreement with that of the exact solution given in Fig-
ure 16-6, the values and signs of the moments at critical sec-
tions agree quite well.
There is one aspect of preliminary design that has not been
discussed that is essential in the handling of the stepped- and
double-column conditions. The non-prismatic nature of
these column types requires input of the moment of inertia
of the upper and lower segments of the column, which, of
course, are not known initially. Therefore, some guidelines
and/or methods are required to obtain reasonable values for
I1 and I2.

(a) no-sway computer model (b) results of no-sway model
Fig. 16-12. Stepped column modeling.
Fig. 16-13. Simplified structure.

l
−
=
2
2
1
2
1
3
2
Pe h
( )
d
H
h
l
3
1
d
l
3
1
d
l
3
2
⎛
⎝
⎜ − +
⎞
⎠
⎟
l
−
=
2
2
2
6
1
3
2
Pe h
( )
d
H
h
l
3
1
d
l
3
1
d
l
3
2
⎛
⎝
⎜ − +
⎞
⎠
⎟
l
− +
=
2
3 3
2
4
1
2
2h d
P ( )
3dh
H
h
l
3
1
d
l
3
1
d
l
3
2
⎛
⎝
⎜ − +
⎞
⎠
⎟
Fig. 16-14. Equations for simplified structure.
Fig. 16-15. Column moments using
the equations in Figure 16-14.
Obtaining Trial Moments of Inertia for Stepped Columns
The upper segment of the stepped column may be sized by
choosing a column section based on the axial load acting on
the upper column portion. Use the appropriate unsupported
length of the column in its weak direction, and determine a
suitable column from the column tables in theAISC Manual.
Select a column about three sizes (by weight) larger to
account for the bending in the upper shaft.
The size of the lower segment of the stepped column
may be obtained by assuming that the gravity load from the
crane is a concentric load applied to one flange (or flange-
channel combination). Experience has shown that a prelim-
inary selection may be made by choosing a member such
that Areq’d = P2/0.40Fy (LRFD) or Areq’d = P2/0.25Fy (ASD),
where Areq’d is the area of one flange or flange plus chan-
nel combination. The depth of the lower shaft is typically
determined by the crane clearance requirements, as shown
in Figure 16-16.
16.2.2 Double Columns (Laced or Tied)
The building column portion of a double column can again
best be selected based on the axial load in the building col-
umn. Select the size of the crane column based on the crane
gravity load applied to the “separate” crane column. The
allowable stress of this portion will typically be based on
the major axis of the column, assuming that the column is
laced or battened to the building column to provide support
about the weak axis. The actual size of the column should be
increased slightly to account for the bending moments. The
moment of inertia of the combined section can be calculated
using standard formulas for geometric properties of built-up
cross sections. If the moment of inertia of the combined sec-
tion is obtained by assuming composite behavior, the lacing
or batten plates connecting the two column sections must be
designed and detailed accordingly.
16.2.3 Single Columns (Bracketed)
The sizes of bracketed columns are often controlled by
wind; therefore, the design should first be made for wind and
subsequently checked for wind plus crane. The preliminary
design procedure for wind or seismic loads can be made by
assuming an inflection point and selecting the preliminary
column size to control sway under wind loads as shown in
Figure 16-17. The approximate procedure is shown in the
bracketed crane column design example in the next section.

Fig. 16-16. Column clearance requirement.
AIST TR-13 recommends that bracket vertical loads
should be limited to 50 kips.
16.3 FINAL DESIGN PROCEDURES
After obtaining the final forces and moments, the final
design of the columns can be performed. The design of a
crane column is unique in that the column has both a varying
axial load and a concentrated moment at the location of the
bracket or “step” in the column.
The best approach for prismatic bracketed columns or for
stepped columns is to design the upper and lower portions
of the columns as individual segments, with the top portion
designed for P1 and the associated upper column moments,
and the lower portion designed for P1 + P2 and the lower col-
umn moments, as shown in Figure 16-18. The column can
be considered laterally braced about the y-axis at the crane
girder elevation.
AIST TR-13 recommends that the design of mill build-
ings be done in accordance with the AISC Specification
provisions. The AISC Specification requires a second-order
analysis to determine the forces and moments in the col-
umns. With proper modeling and analysis of the structure,
the complication of determining the effective length of the
Fig. 16-17. Approximate sway calculation.
columns is eliminated. The effective length, Kx, for the col-
umn sections can be taken as 1.0, and moment magnifiers to
take care of the P-δ requirements need not be used if the col-
umn sections are modeled with node points along the length
of each column section.
Axial Compressive Strength
The majority of columns in mill buildings are nonslender,
and therefore the nominal compressive strength, Pn, is deter-
mined based on the limit state of flexural buckling using
AISC Specification Section E3.
Flexural Strength
For simple bending, the member is loaded in a plane parallel
to a principal axis that passes through the shear center or is
restrained against twisting at load points and supports. The
nominal flexural strength, Mn, is the lower value obtained
according to the limit states of yielding and lateral-torsional
buckling using AISC Specification Section F2.
Combined Axial Force and Flexure
The interaction of required strengths for members subject to
combined axial forces and flexure must satisfy the interac-
tion equations of AISC Specification Chapter H.
AISC Design Aids
AISC Manual Table 6-2 provides available strengths of
ASTM A992 W-shapes and may be of use when designing
members with combined effects.
The following examples illustrate the column design
procedure.

Fig. 16-18. Column loads.

Example 16.3.1—Bracketed Crane Column Design Example
Given:
Design the column shown in Figure 16-19 using ASTM A992 steel. The column loading is as follows:
D = 20 psf
Lr = 30 psf
Cds = 20 kips (left-side column)
Cvs = 30 kips (left-side column)
Cds = 20 kips (right-side column)
Cvs = 2 kips (right-side column)
Css = 3 kips
Solution:
Use the following AIST TR-13 load combinations:
LRFD: D C C C L
1.2 1.6 0.5
ds vs ss r
( ) ( )
+ + + +
ASD: D C C C L
0.75
ds vs ss r
( )
+ + + +
where
D = dead load
Cds =
crane dead load for a single crane with crane trolley positioned to produce the maximum load effect for the element in
consideration; crane dead load includes weight of the crane bridge, end trucks, and trolley
Css = crane side thrust from a single crane
Fig. 16-19. Bracketed crane column example.

Cvs =
crane lifted load for a single crane with crane trolley positioned to produce the maximum load effect for the element
under consideration
Lr = roof live load
Nominal nodal loads on the truss:
P 20 psf 20 ft 5 ft 1 kip/1,000 lb
2.00 kips
D ( )
( )( )( )
=
=
P 30 psf 20 ft 5 ft 1 kip/1,000 lb
3.00 kips
L ( )
( )( )( )
=
=
Wind loads at eave:
P 9.00 kips
w =
Ten year wind:
P 6.75 kips
w10 =
The column is laterally braced about the y-axis at the truss top chord and bottom chord and 16 ft above the floor.
Preliminary Design
Because this structure is quite tall, it is likely that lateral sway movement may control the column size. Thus, it is recommended
that the preliminary design of the column be based on deflection considerations.
Base the allowable sway at the rail height on the minimum of H/240 or 1.0 in. Use a 10-year wind and/or the crane lateral load
as the load criterion. The rail height is assumed to be 24 in. above the bracket.
For the wind load:
H
240
45 ft 12 in./ft
240
2.25 in.
( )( )
=
=
For a fixed-fixed column with Pw10 = 6.75 kips, the eave deflection is approximately:
H
EI
24
3
=
Δ
10
w
P

(16-1)
Assuming Pw10 is divided equally between the windward and leeward columns:
I
6.75 kips 2 45 ft 1,728 in. /ft
24 29,000 ksi
764 in.
x
3 3 3
4
(
( )( )
( )( )(1.0 in.)
=
=
Try a W16×77. From AISC Manual Table 1-1:
Ix = 1,110 in.4
764 in.4
o.k.
The fixed-base frame model is shown in Figure 16-20. The vertical crane loads are located at nodes N27 and N29. The lateral
crane loads are applied at nodes N28 and N30.

Member Properties
The following member sizes are used for columns and truss members:
Columns: W16×77
Truss top chord: WT7×34
Truss bottom chord: WT7×21.5
Truss web members: 2L3×3×c
Based on the model, the eave deflection determined from a first-order analysis due to a 10-year wind is 1.21 in. The deflection
of the rail can then be determined:
34 ft
45 ft
1.21 in.
0.914 in.
rail ( )
=
Δ ⎛
⎝
⎞
⎠
=
Determine the deflection at the crane rail due to the crane vertical and horizontal loads using the model.
The model loads are:
Node N27:
P2 = 50 kips
M = −1,000 kip-in. (based on an eccentricity of 8 in. + 12 in. = 20 in.)
Node N29:
P2 = 10 kips
M = 200 kip-in. (based on an eccentricity of 8 in. + 12 in. = 20 in.)
Nodes N28 and N30:
PH = 3 kips
P1 is the summation of the roof loads (D + Lr) to each column. Design nodal loads are shown in Table 16-1.
Results indicate a deflection of 1.13 in. at the rail height.
Fig. 16-20. Frame model.

Table 16-1. Design Nodal Loads
Nodes LRFD ASD
N6-N17
P 1.2 2.00 kips 0.5 3.00 kips
3.90 kips
y ( ) ( )
= − + −
= −
P 2.00 kips 0.75 3.00 kips
4.25 kips
y ( )
= − + −
= −
N2 and N14
P
3.90 kips
2
1.95 kips
y =
−
= −
P
4.25 kips
2
2.13 kips
y =
−
= −
N27
P 1.2 20.0 kips 1.6 30.0 kips
72.0 kips
y ( ) ( )
= − + −
= −
P 20.0 kips 0.75 30.0 kips
42.5 kips
y ( )
= − + −
= −
N27
M 72.0 kips 20.0 in.
1
,440 kip-in.
( )( )
= −
= −
M 42.5 kips 20.0 in.
850 kip-in.
( )( )
= −
= −
N29
P 1.2 20.0 kips 1.6 2.00 kips
27.2 kips
y ( ) ( )
= − + −
= −
P 20.0 kips 0.75 −2.00 kips
21.5 kips
y ( )
= − +
= −
N29
M 20.0 in. 27.2 kips
544 kip-in.
( )
( )
=
=
M 20.0 in. 21.5 kips
430 kip-in.
( )
( )
=
=
N28
P 1.6 3.00 kips
4.80 kips
H ( )
=
=
P 0.75 3.00 kips
2.25 kips
H ( )
=
=
N30
P 1.6 3.00 kips
4.80 kips
H ( )
=
=
P 0.75 3.00 kips
2.25 kips
H ( )
=
=
Check the Column Available Strength
A second-order elastic analysis is performed using the requirements of AISC Specification Chapter C. For ASD the loads must be
multiplied by 1.6, the analysis performed, and the results divided by 1.6 to obtain the ASD results. The moments based on LRFD
are shown in Figure 16-21(a). The results for ASD are shown in Figure 16-21(b).
By observation the lower shaft has higher force demands than the upper shaft; thus, the lower shaft will be checked.
(a) LRFD (b) ASD
Fig. 16-21. Moment diagrams at right column.

For the fixed-base condition of the column, the effective length, Lc, is:
Lc = KL
= 0.5(32 ft)
= 16 ft
LRFD ASD
From the analysis:
P
M
56.6 kips
137 kip-ft
u
u
=
=
From AISC Manual Table 6-2, for a W16×77 with
Lc = Lb = 16 ft:
P
M
654 kips
482 kip-ft
c n
b nx
ϕ =
ϕ =
From AISC Specification Section H1.1:
P
P
P
P
56.6 kips
654 kips
0.0865
r
c
u
n
=
ϕ
=
=
Because Pr/Pc 0.2, use AISC Specification
Equation H1-1b:
P
P
M
M
M
M
2
1.0
0.0865
2
137 kip-ft
482 kip-ft
0 0.327 1.0
r
c
rx
cx
ry
cy
o.k.
+ +
⎛
⎝
⎜
⎞
⎠
⎟ ≤
+ +
⎛
⎝
⎜
⎞
⎠
⎟ =
From the analysis:
P
M
52.0 kips
91.0 kip-ft
a
a
=
=
Lc = Lb = 16 ft:
P
M
435 kips
321 kip-ft
n c
nx b
=
Ω
=
Ω
P
P Ω
P
P
52.0 kips
435 kips
0.120
r
c
a
n
=
=
=
Equation H1-1b:
P
P
M
M
M
M
2
1.0
0.120
2
91.0 kip-ft
321 kip-ft
0 0.343 1.0
r
c
rx
cx
ry
cy
o.k.
+ +
⎛
⎝
⎜
⎞
⎠
⎟ ≤
+ +
⎛
⎝
⎜
⎞
⎠
⎟ =
Because the member is adequate using Cb =1.0, there is no need to refine the calculations for a calculated Cb.
Example 16.3.2—Stepped Crane Column Design Example
Given:
Design the column shown in Figure 16-22 using ASTM A992 steel. The column loading is as follows:
D = 20 psf
Lr = 30 psf
Cds = 20 kips (left-side column)
Cvs = 30 kips (left-side column)
Cds = 20 kips (right-side column)
Cvs = 2 kips (right-side column)
Css = 3 kips
Solution:
Use the following AIST TR-13 load combinations:
LRFD: D C C C L
1.2 1.6 0.5
ds vs ss r
( (
) )
+ + + +
ASD: D C C C L
0.75
ds vs ss r
( )
+ + + +

Nominal nodal loads on the truss:
P 20 psf 20 ft 5 ft 1 kip/1,000 lb
2.00 kips
D ( )
( )( )( )
=
=
P 30 psf 20 ft 5 ft 1 kip/1,000 lb
3.00 kips
L ( )
( )( )( )
=
=
Wind loads at eave:
Pw = 9.00 kips
Ten-year wind:
Pw10 = 6.75 kips
The column is laterally braced about the y-axis at the truss top chord and bottom chord and 16 ft above the floor.
Preliminary Design
The crane load axial force is 50 kips (unfactored). For the top column section, the unbraced length is 8 ft.
Estimate the flange area based on the crane vertical load.
P
A
P
F
50.0 kips
0.25
(ASD)
=
50.0 kips
0.25 50 ksi
4.00 in.
req’d
y
2
2
2
( )
=
=
=
Fig. 16-22. Stepped crane column example.

Estimate the section for the top column to be a W12×65.
A W24 section is required for the bottom column section for crane clearance; try a W24×68.
From AISC Manual Table 1-1:
bf = 8.97 in.
tf = 0.585 in.
A b t
8.97 in. 0.585 in.
5.25 in.
flange f f
2
( )( )
=
=
=
The fixed-base frame model is shown in Figure 16-23. The vertical crane loads are located at nodes N27 and N29. The lateral
crane loads are applied at nodes N28 and N30.
Trial Member Properties
The following member sizes are used for columns and truss members:
Upper-shaft columns: W12×65
Lower-shaft column: W24×68
Truss top chord: WT7×34
Truss bottom chord: WT7×21.5
Truss web members: 2L3×3×c
The model loads are:
Node N27:
P2 = 50 kips
M = −600 kip-in. (based on an eccentricity of 12 in.)
Node N29:
P2 = 10 kips
M = 120 kip-in. (based on an eccentricity of 12 in.)
Fig. 16-23. Frame model.

Nodes N28 and N30:
PH = 3 kips
P1 is the summation of the roof loads (D + Lr) to each column. Design nodal loads are shown in Table 16-2.
Ignore the eccentricity of the upper shaft on the lower shaft (e = 6 in.)
Determine the deflection at the crane rail due to the crane vertical and lateral crane loads using the model.
Based on the model, the eave deflection from a first-order analysis due to the 10-year wind is 1.16 in.
Results indicate a deflection of 0.98 in. at the rail height. o.k.
Check the Column Available Strength
A second-order elastic analysis is performed using the requirements of AISC Specification Chapter C. For ASD the loads must be
multiplied by 1.6, the analysis performed, and the results divided by 1.6 to obtain the ASD results. The moments based on LRFD
are shown in Figure 16-24(a). The results for ASD are shown in Figure 16-24(b).
Lower Shaft—W24×68
For the fixed-base condition of the column, the effective length, Lc, is:
Lc = KL
= 0.5(32 ft)
= 16 ft
Table 16-2. Design Nodal Loads
Nodes LRFD ASD
N6–N17
P 1.2 −2.00 kips 0.5 −3.00 kips
3.90 kips
y ( ) ( )
= +
= −
P −2.00 kips 0.75 −3.00 kips
4.25 kips
y ( )
= +
= −
N2 and N4
P
3.90 kips
2
1.95 kips
y =
−
= −
P
4.25 kips
2
2.13 kips
y =
−
= −
N27
P 1.2 −20.0 kips 1.6 −30.0 kips
72.0 kips
y ( ) ( )
= +
= −
P −20.0 kips 0.75 −30.0 kips
42.5 kips
y ( )
= +
= −
N27
M 20.0 in. −72.0 kips
1,440 kip-in.
( )
( )
=
= −
M 20.0 in. −42.5 kips
850 kip-in.
( )
( )
=
= −
N29
P 1.2 −20.0 kips 1.6 −2.00 kips
27.2 kips
y ( ) ( )
= +
= −
P −20.0 kips 0.75 −2.0 kips
21.5 kips
y ( )
= +
= −
N29
M 20.0 in. 27.2 kips
544 kip-in.
( )
( )
=
=
M 20.0 in. 21.5 kips
430 kip-in.
( )
( )
=
=
N28
P 1.6 3.00 kips
4.80 kips
H ( )
=
=
P 0.75 3.00 kips
2.25 kips
H ( )
=
=
N30
P 1.6 3.00 kips
4.80 kips
H ( )
=
=
P 0.75 3.00 kips
2.25 kips
H ( )
=
=

LRFD ASD
From the analysis:
P
M
56.0 kips
215 kip-ft
u
u
=
=
Lc = Lb = 16 ft:
P
M
418 kips
465 kip-ft
c n
b nx
ϕ =
ϕ =
P
P
P
P
56.0 kips
418 kips
0.134
r
c
u
n
=
ϕ
=
=
Equation H1-1b:
P
P
M
M
M
M
2
1.0
0.134
2
215 kip-ft
465 kip-ft
0 0.529 1.0
r
c
rx
cx
ry
cy
o.k.
+ +
⎛
⎝
⎜
⎞
⎠
⎟ ≤
+ +
⎛
⎝
⎜
⎞
⎠
⎟ =
From the analysis:
P
M
52.1 kips
128 kip-ft
a
a
=
=
Lc = Lb = 16 ft:
P
M
278 kips
309 kip-ft
n c
nx b
=
Ω
=
Ω
P
P
P
P Ω
52.1 kips
278 kips
0.187
r
c
a
n
=
=
=
Equation H1-1b:
P
P
M
M
M
M
2
1.0
0.187
2
128 kip-ft
309 kip-ft
0 0.508 1.0
r
c
rx
cx
ry
cy
o.k.
+ +
⎛
⎝
⎜
⎞
⎠
⎟ ≤
+ +
⎛
⎝
⎜
⎞
⎠
⎟ =
(a) LRFD (b) ASD
Fig 16-24. Moment diagrams at right column.

Because the member is adequate using Cb =1.0, there is no need to refine the calculations for a calculated Cb.
Upper Shaft—W12×65
For the pinned-base condition of the column, the effective length, Lc, is:
Lc = KL
= 1.0(8 ft)
= 8 ft
LRFD ASD
From the analysis:
P
M
29.0 kips
97.0 kip-ft
u
u
=
=
Lc = Lb = 8 ft:
P
M
798 kips
356 kip-ft
c n
b nx
ϕ =
ϕ =
P
P
P
P
29.0 kips
798 kips
0.0363
r
c
u
n
=
ϕ
=
=
Because Pr/Pc 0.2, use AISC Specification Equation
H1-1b:
P
P
M
M
M
M
2
1.0
0.0363
2
97.0 kip-ft
356 kip-ft
0 0.291 1.0
r
c
rx
cx
ry
cy
o.k.
+ +
⎛
⎝
⎜
⎞
⎠
⎟ ≤
+ +
⎛
⎝
⎜
⎞
⎠
⎟ =
From the analysis:
P
M
30.6 kips
68.5 kip-ft
a
a
=
=
Lc = Lb = 8 ft:
P
M
531 kips
237 kip-ft
n c
nx b
Ω =
Ω =
P
P
P
P
30.6 kips
531 kips
0.0576
r
c
a
n
=
Ω
=
=
Because Pr/Pc 0.2, use AISC Specification Equation
H1-1b:
P
P
M
M
M
M
2
1.0
0.0576
2
68.5 kip-ft
237 kip-ft
0 0.318 1.0 o.k.
r
c
rx
cx
ry
cy
+ +
⎛
⎝
⎜
⎞
⎠
⎟ ≤
+ +
⎛
⎝
⎜
⎞
⎠
⎟ =
16.4 ECONOMIC CONSIDERATIONS
Although it is not possible to provide a clear-cut rule of
thumb as to the most economical application of the vari-
ous crane columns—that is, bracketed, stepped, or separate
crane column, due to differences in shop techniques—it is
possible to generalize them to some degree.
The stepped column will be economical if it is clean—that
is, fabricated without a face channel or extra welded attach-
ments, as shown in Figure 16-25. In fact, for many jobs, a
clean stepped column can prove economical as compared
to the bracketed column, even for light loads. Also, the cap
plate can be made thick enough to eliminate the need for a
stiffener under the upper shaft’s interior flange.
Separate crane columns are economical for heavy cranes.
Fabricators favor tying the crane column to the building col-
umn with short W-shapes acting as a diaphragm as opposed
to a lacing system using angles, as shown in Figure 16-26.
Lacing systems are economical as compared to the dia-
phragm system if miscellaneous framing pieces are not
required. For example, if the building column flange width
is equal to the crane column depth, the columns can be laced
economically using facing angles, as shown in Figure 16-27.
Bracketed columns are generally most efficient up to
bracket loads of 25 kips. Crane reactions between 25 kips
and 50 kips may best be handled by either a bracketed col-
umn or a stepped column.
If the area of one flange of a stepped column multiplied
by 0.5Fy is less than the crane load on the column, a separate
crane column should definitely be considered.

Fig. 16-25. “Clean” column. Fig. 16-26. Connections with W-shapes.
Fig.16-27. Laced columns (box plates not shown).

Chapter 17
Other Crane Considerations
17.1 OUTSIDE CRANES
Outside cranes are common in many factories for scrap han-
dling, parts handling, and numerous other operations. There
are several important aspects of outside crane usage that are
unique to that type of crane.
1. The exterior exposure in many climates requires that
extra attention be given to painting and general mainte-
nance, material thickness, and the elimination of pockets
that would collect moisture.
2. Due to drive aisles, railways, and other similar restric-
tions, exterior cranes often require longer spans than
interior cranes. The outside crane has no building col-
umns from which to derive lateral support. Therefore,
long, unbraced spans are more common in these instal-
lations. Horizontal bracing trusses, wide-truss columns,
or other bracing elements must often be employed to
achieve stability.
3. Long spans may dictate that trusses, rather than plate
girders or rolled sections, be used for the runway beams.
This can have certain advantages, including improved
stiffness. The disadvantages are clearly the increased
depth plus joints that are highly susceptible to fatigue
problems. Secondary stresses must be calculated and
included in the fatigue analysis for trusses used as crane
girders.
4. Another special girder that may be appropriate for use
in these long-span applications is the trussed girder. This
“hybrid” involves the coupling of a girder (top flange)
and a truss. The member can develop excellent stiffness
characteristics and many times can temporarily support
the crane weight even if a truss member is damaged.
As with the basic truss, the overall greater depth is a
disadvantage.
5. Still another solution to the long-span problem may lie
in the use of “box” or “semi-box” girders. An excellent
reference on this subject was developed by Schlenker
(1972). These girders have excellent lateral and torsional
strength. In addition, the problem associated with off-
center crane rails is eliminated; however, internal dia-
phragm plates are generally required to keep the box
square during fabrication and will reduce fatigue life if
not properly designed for fatigue.
6. Brittle fracture should be considered for cranes operat-
ing in low-temperature environments.
17.2 UNDERHUNG CRANES
Underhung cranes in industrial buildings are very common
and quite often prove to be economical for special applica-
tions. Underhung cranes are characterized by the fact that
the end trucks and the bridge are supported from the struc-
ture above. Underhung cranes are generally used for less
severe applications and for lighter loads as compared to
overhead cranes. One of the distinct operational advantages
that underhung cranes possess is that they can be arranged
to provide for trolley transfer from one runway or aisle to
another. Proper provision in the design must be made for
handling lateral and impact loads from underhung cranes.
The concepts presented in this Guide regarding load trans-
fer are, in general, applicable to underhung crane systems.
Because these cranes are generally supported by roof mem-
bers, load is not transferred directly to the columns, and
therefore, the column design does not involve the moment
distribution problems of the top-running crane column. Par-
ticular attention should be paid to the method of hanging
the cranes. Fatigue problems with these connections have
existed in the past, and proper provisions must be made with
the hanging connection to guarantee adequate service life.
Hanger systems should provide for vertical adjustment to
properly adjust the elevation of the runway beam. After the
runways are positioned vertically, a lateral anti-sway brace
should be attached. The sway brace prevents the hanger sys-
tem from flexing perpendicular to the runway. Lateral braces
should only be provided on one side of the hanger system
as shown in Figure 17-1. This keeps the crane in alignment
and prevents lateral forces from being generated on the han-
gar system as the crane travels up and down on the runway.
Most hanger systems fatigue at a relatively low stress level if
they are allowed to sway. In addition to the lateral anti-sway
braces, longitudinal braces should be installed parallel to the
runway beams to prevent sway along the length of the run-
way. These braces should be placed at approximately 100-ft
intervals and at all turns in the runway.
Runway splices can be accomplished in many ways. The
splice should allow for a smooth-running crane as the wheels
transfer from one beam to the next. A typical splice detail is
shown in Figure 17-2.
Especially for higher usage underhung cranes, periodic
inspection of the bottom flange must be undertaken to ensure
that the wear of the running surface does not compromise the
structural integrity of the wheel support.
Many crane suppliers prefer to supply the runway beams.
The building designer must carefully coordinate hanger

locations and hanger reactions with the crane supplier. Many
times, the structure must be designed prior to the selection
of the crane system, and the hanger locations and reactions
must be estimated by the building designer. Hanger reactions
can be calculated from manufacturer’s catalogs. Hangers
should be provided at a 15- to 20-ft spacing if possible. The
deflection limit for underhung crane runway beams due to
wheel loads should be limited to span divided by 450.
In addition to the various AISC Specification checks that
must be made for the design of underhung crane beams, a
bottom flange localized combined stress check must be
made to determine the effects of the wheel contact load on
the bottom flange. The effect of the concentrated wheel load
can be to “cold work” the steel in the bottom flange, which
in the long term can result in autofrettage, cracking, and
break-off of portions of the bottom flange. Contained in the
CMAA Specification for Top Running and Under Running
Single Girder Electric Overhead Cranes Utilizing Under
Running Trolley Hoists—No. 74 (CMAA, 2015b) is a sug-
gested design approach for the examination of the wheel
contact stresses.
Monorail cranes are similar in nature to underhung cranes
except they are used to transport materials in prescribed
paths.
If open web steel joists are used to support the hangers,
special joists must be designated at the support hangers. The
manufacturer could be asked to mark the special joists to
avoid confusion in the field between the special crane sup-
port joists and the typical joists.
The crane beam and monorail support hangers must
load the joist at a panel point; otherwise, concentrated load
reinforcement must be provided or the manufacturer must
design the joist chord for the induced bending. The hang-
ers should allow for vertical adjustment. This will allow the
crane beams to be leveled after the roofing has been applied
and the dead load deflection of the roof system has occurred.
The vertical adjustability of the hangers will also accommo-
date the differences in elevation caused by fabrication and
erection tolerances.
Care should be taken in the design and detailing of the
lateral braces. The brace is intended to resist lateral load;
however, the brace may inadvertently pick up some of the
Fig. 17-1. Underhung crane support system.
Fig. 17-2. Underhung crane beam splice.

vertical load depending on its stiffness relative to the vertical
hangers. Because the hangers and the lateral brace are not
located precisely at a panel point, their loads and locations
must be supplied to the joist manufacturer.
If the crane runway support beams are parallel to the
joists, the lateral brace will have to extend to the top chord
of an adjacent joist, and horizontal members will have to
be added directly under the deck to transfer the thrust load
into the roof deck. A typical hanger and brace for this situa-
tion are illustrated in Figure 17-3. When the runway support
beams are perpendicular to the joist, a detail similar to that
shown in Figure 17-4 may be appropriate.
The tractive longitudinal force at each runway is typi-
cally specified as 10% of the total maximum wheel loads
supported by that side of the runway. The longitudinal force
created by the crane hitting the crane stops may exceed the
tractive longitudinal force. The stopping force is a function
of the crane travel speed and the length of stroke of the crane
bumper. This bumper force can be controlled by the selec-
tion of the bumper. The resulting load to the support system
should be coordinated between the engineer and the crane
supplier. A bracing system is required to resist the longitu-
dinal crane force. If the crane runway runs parallel to the
joists, the longitudinal thrusts are transferred through the
joist diagonals to the top chord and into the roof deck. The
typical hanger detail will require modification to also trans-
fer the longitudinal load into the joist.
Clamp-type hangers may be used to attach hangers to the
bottom chord of joists. However, the engineer must design
the clamps to avoid bending the outstanding legs of the joist
Fig. 17-3. Runway support beams parallel to joists.
Fig. 17-4. Typical hanger at joist.

can result if proper maintenance is not provided. Crane
rails must also be inspected for uneven bearing to minimize
fatigue problems.
If fatigue cracks occur and must be repaired, the repair
procedure may create additional problems if proper proce-
dures are not taken. Simple welding of doubler plates, stiff-
eners, or other reinforcement may create a “notch effect”
that could be more serious than the original problem. Engi-
neers should use common sense in detailing procedures for
repair of fatigue cracks. In particular, they should not worsen
the fatigue problem with the repair. Referral to AISC Specifi-
cation Appendix 3 is essential.
chord. Clamps and hangers are not part of the components
designed and supplied by the joist manufacturer.
Depending on the trolley location, either the left or right
hanger load may be larger. Given the shifting shear and
moment diaphragm created by the possible crane loading
conditions, the use of KCS series (constant-shear) joists
should be considered for this situation.
17.3 MAINTENANCE AND REPAIR
As stated earlier, crane buildings require an extra measure
of maintenance. Crane rail alignment is especially critical.
Wear on the crane and rail and potential fatigue problems

Chapter 18
Summary and Design Procedures
Many concepts have been presented in this Guide relative to
the design and analysis of structural frames for crane build-
ings. In an effort to optimize design time, the following pro-
cedural outline has been developed for the designer:
1. Determine the best geometrical layout for the building
in question.
2. Design the crane girders and determine column and
frame forces from the crane loads.
3. Perform preliminary design of the crane columns.
4. Design the roof trusses or roof beams for dead loads
and live loads.
5. Determine all loading conditions for which the entire
frame must be analyzed.
6. Analyze the frame in question for dead, live, wind,
and seismic loads. This analysis should be performed
without load sharing from the adjacent frames and
determine the lateral stiffness of the frame.
7. Analyze the frame (considering load sharing) for
crane loads when horizontal trusses are used.
8. Combine moments and forces from the two analyses
for subsequent design.
9. Perform the final design of columns, trusses, braces,
and details.

Appendix
Table A-1. W-Shapes with Cap Channels: Section Properties
W-Shape Channel
Total
Wt.,
lb/ft
Axis X-X Axis Y-Y
Ix, in.4
S1, in.3
S2, in.3
y1, in. Zx, in.3
It, in.4
Syt, in.3
Zyt, in.3
W36×150 MC18×42.7 193 12000 553 831 21.8 738 689 76.6 109
C15×33.9 184 11500 546 764 21.1 716 450 60.1 84.6
W33×141 MC18×42.7 184 10000 490 750 20.4 652 676 75.1 106.8
C15×33.9 175 9580 484 689 19.8 635 437 58.2 82.5
W33×118 MC18×42.7 161 8280 400 656 20.7 544 648 72.0 99.6
C15×33.9 152 7900 395 596 20.0 529 409 54.5 75.3
W30×116 MC18×42.7 159 6900 365 598 18.9 492 636 70.7 98.5
C15×33.9 150 6590 360 544 18.3 480 397 52.9 74.2
W30×99 MC18×42.7 142 5830 304 533 19.2 412 619 68.7 93.6
C15×33.9 133 5550 300 481 18.5 408 380 50.6 69.3
W27×94 C15×33.9 128 4530 268 435 16.9 357 377 50.3 69.4
W27×84 C15×33.9 118 4050 237 403 17.1 316 368 49.0 66.7
W24×84 C15×33.9 118 3340 217 367 15.4 286 362 48.3 66.5
C12×20.7 105 3030 211 302 14.3 275 176 29.4 41.3
W24×68 C15×33.9 102 2710 173 321 15.7 232 350 46.7 62.6
C12×20.7 88.7 2440 168 258 14.5 224 164 27.4 37.4
W21×68 C15×33.9 102 2180 156 287 13.9 207 347 46.3 62.5
C12×20.7 88.7 1970 152 232 12.9 200 161 26.9 37.3
W21×62 C15×33.9 95.9 2000 142 272 14.1 189 344 45.8 61.2
C12×20.7 82.7 1800 138 218 13.0 183 158 26.3 36.0
W18×50 C15×33.9 83.9 1250 100 211 12.5 133 335 44.7 58.8
C12×20.7 70.7 1120 97.3 166 11.5 127 149 24.8 33.6
W16×36 C15×33.9 69.9 748 64.5 160 11.6 86.8 327 43.6 56.1
C12×20.7 56.7 670 62.8 123 10.7 83.2 141 23.5 30.9
W14×30 C12×20.7 50.7 447 46.7 98.1 9.57 62.0 139 23.1 30.0
C10×15.3 45.3 420 46.0 84.5 9.11 60.3 77.1 15.4 20.3
W12×26 C12×20.7 46.7 318 36.8 82.1 8.63 48.2 138 22.9 29.6
C10×15.3 41.3 299 36.3 70.5 8.22 47.0 76.0 15.2 19.9
y1 = distance from the bottom of the tension flange to the elastic neutral axis, in.

Table A-2. W-Shapes with Cap Channels: Lateral-Torsional Buckling Properties
W Shape Channel rt, in. ho, in. FL, ksi Lp, in. Lr, in.
W36×150 MC18×42.7 5.09 35.0 33.3 135 503
C15×33.9 4.32 35.0 35.0 115 417
W33×141 MC18×42.7 5.09 32.4 32.7 135 514
C15×33.9 4.31 32.4 35.0 114 420
W33×118 MC18×42.7 5.28 32.1 30.5 140 541
C15×33.9 4.45 32.2 33.1 118 436
W30×116 MC18×42.7 5.21 29.2 30.5 138 545
C15×33.9 4.36 29.1 33.1 115 438
W30×99 MC18×42.7 5.39 29.3 28.5 143 573
C15×33.9 4.50 29.0 31.2 119 457
W27×94 C15×33.9 4.47 26.1 30.8 118 465
W27×84 C15×33.9 4.56 26.0 29.4 121 481
W24×84 C15×33.9 4.47 23.3 29.6 118 486
C12×20.7 3.49 23.3 34.9 93.0 346
W24×68 C15×33.9 4.65 23.1 26.9 123 518
C12×20.7 3.62 23.0 32.6 96.0 361
W21×68 C15×33.9 4.58 20.4 27.2 121 529
C12×20.7 3.55 20.4 32.8 94.0 366
W21×62 C15×33.9 4.66 20.3 26.1 123 543
C12×20.7 3.61 20.3 31.7 96.0 373
W18×50 C15×33.9 4.76 17.5 23.7 126 636
C12×20.7 3.67 17.5 29.3 97.0 427
W16×36 C15×33.9 4.95 15.3 20.2 131 700
C12×20.7 3.85 15.3 25.5 102 451
W14×30 C12×20.7 3.92 13.3 23.8 104 491
C10×15.3 3.21 13.3 27.2 85.0 366
W12×26 C12×20.7 3.96 11.7 22.4 105 537
C10×15.3 3.24 11.7 25.7 86.0 395
FL calculated using AISC Specification Equation F4-6a/F4-6b.
Lp calculated using AISC Specification Equation F4-7.
Lr calculated using AISC Specification Equation F4-8.

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