Simplified design of reinforced concrete buildings

Simplified
Design of
Reinforced
Concrete
Buildings
Fourth Edition
Mahmoud E. Kamara
Lawrence C. Novak
An organization of cement companies to improve and
extend the uses of portland cement and concrete
through market development, engineering, research,
education, and public affairs work.
5420 Old Orchard Road, Skokie, Illinois 60077-1083
www.cement.org
ENGINEERING BULLETIN EB204

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© 2011 Portland Cement Association
Fourth edition
First printing 2011
Library of Congress Catalog Card Number 93-30929
ISBN 978-0-89312-273-7
This publication was prepared by the Portland Cement Association for the purpose of suggesting
possible ways of reducing design time in applying the provisions contained in the ACI318-11
Building Code Requirements for Structural Concrete.
Simplified design procedures stated and illustrated throughout this publication are subject to
limitations of applicability. While such limitations of applicability are, to a significant extent, set
forth in the text of this publiction, no attempt has been made to state each and every possible lim-
itation of applicability. Therefore, this publication is intended for use by professional personnel
who are competent to evaluate the information presented herein and who are willing to accept
responsibility for its proper application.
Portland Cement Association (“PCA”) is a not-for-profit organization and provides this publi-
cation solely for the continuing education of qualified professionals. THIS PUBLICATION
SHOULD ONLY BE USED BY QUALIFIED PROFESSIONALS who possess all required
license(s), who are competent to evaluate the significance and limitations of the information
provided herein, and who accept total responsibility for the application of this information.
OTHER READERS SHOULD OBTAIN ASSISTANCE FROM A QUALIFIED PROFES-
SIONAL BEFORE PROCEEDING.
PCA AND ITS MEMBERS MAKE NO EXPRESS OR IMPLIED WARRANTY WITH
RESPECT TO THIS PUBLICATION OR ANY INFORMATION CONTAINED HEREIN. IN
PARTICULAR, NO WARRANTY IS MADE OF MERCHANTABILITY OR FITNESS FOR
A PARTICULAR PURPOSE. PCA AND ITS MEMBERS DISCLAIM ANY PRODUCT
LIABILITY (INCLUDING WITHOUT LIMITATION ANY STRICT LIABILITY IN TORT)
IN CONNECTION WITH THIS PUBLICATION OR ANY INFORMATION CONTAINED
HEREIN.

Foreword
The Building Code Requirements for Structural Concrete (ACI 318) is an authoritative document often adopted and
referenced as a design and construction standard in building codes around the country as well as in the
specifications of several federal agencies, its provisions thus becoming law. Whether ACI 318 is enforced as part
of building regulations or is otherwise utilized as a voluntary consensus standard, design professionals use this
standard almost exclusively as the basis for the proper design and construction of reinforced concrete buildings.
The ACI 318 standard applies to all types of building uses; structures of all heights ranging from the very
tall high-rise down to single-story buildings; facilities with large areas as well as those of nominal size; buildings
having complex shapes and those primarily designed as uncomplicated boxes; and buildings requiring structurally
intricate or innovative framing systems in contrast to those of more conventional or traditional systems of
construction. The general provisions developed to encompass all these extremes of building design and construc-
tion tend to make the application of ACI 318 complex and time consuming. However, this need not necessarily be
the case, as is demonstrated in the publication.
This book has been written as a timesaving aid for use by experienced professionals who consistently seek ways to
simplify design procedures.
This fourth edition of the book is based onACI 318-11. The seismic and wind load provisions were updated to com-
ply with the International Building Code (2009 IBC) and ASCE7-05. Throughout the first eleven chapters, equa-
tions, design aids, graphs, and code requirements have been updated to the current Codes. New timesaving design
aids were added to expand the use of the book beyond the originally intended 4 ksi concrete making the approach
applicable to a wider range of concrete strengths. Also, expanded illustrations of the theory and fundamentals were
added. A new chapter on sustainable design (Chapter 12) has been included to introduce the key ideas addressed
by today’s green design approaches and ways in which concrete can be used to build sustainably.
In some of the example problems, the results obtained from the simplified design methods are compared to those
obtained from computer programs. These comparisons readily show that the simplified methods yield satisfactory
results within the stated limitations.
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Design professionals reading and working with the material presented in this book are encouraged to send in
their comments to PCA, together with any suggestions for further design simplifications. PCA would also be
grateful to any reader who would bring any errors or inconsistencies to our attention. Any suggestion for
improvement is always genuinely welcome. Any errata to this book or other PCA publications may be found
by checking http://www.cement.org/bookstore/errata.asp
iv

Acknowledgments
The authors acknowledge their indebtedness to the authors and editors of the previous three editions of this book.
Appreciation is due to many colleagues who provided invaluable suggestions.
Thanks go to ASTM and ACI for the use of their material and documents referenced throughout the book.
A special thanks goes to PCA Buildings Committee and all the PCA members; without their continued support,
this publication would not be possible.
Sincere appreciation is due to James A. Farny, PCA Buildings Market Manager, for writing Chapter 12
“Introduction to Sustainable Design”. Thanks to Wes Ikezoe who managed the production of this complex book
including its many tables and figures, his work and patience is greatly appreciated. Thanks are due toArlene Zapata
for the cover design.
Finally the authors wish to express their thanks to all those who in one way or another contributed to the success-
ful completion of this book.
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Table of Contents
Chapter 1—A Simplified Design Approach. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 1-1
1.1 THE BUILDING UNIVERSE. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 1-1
1.2 COST EFFICIENCIES . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 1-2
1.3 THE COMPLEX CODE . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 1-3
1.3.1 Complex Structures Require Complex Designs . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 1-3
1.4 A SIMPLE CODE . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 1-4
1.5 PURPOSE OF SIMPLIFIED DESIGN . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 1-5
1.6 SCOPE OF SIMPLIFIED DESIGN. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 1-5
1.7 BUILDING EXAMPLES . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 1-7
1.7.1 Building No. 1—3-Story Pan Joist Construction . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 1-8
1.7.2 Building No. 2—5-Story Flat Plate Construction. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 1-10
1.8 PRELIMINARY DESIGN . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 1-12
1.8.1 Floor Systems . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 1-12
1.8.2 Columns . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 1-16
1.8.3 Shearwalls. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 1-16
1.8.4 Footings . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 1-20
1.8.5 Fire Resistance . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 1-20
References . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 1-20
Chapter 2—Simplified Frame Analysis. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 2-1
2.1 INTRODUCTION . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 2-1
2.2 LOADING . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 2-1
2.2.1 Service Loads . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 2-1
2.2.2 Wind Loads . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 2-2
2.2.2.1 Example: Calculation of Wind Loads – Building #2 . . . . . . . . . . . . . . . . . . . . . . . . . . . 2-4
2.2.2.2 Example: Calculation of Wind Loads – Building #1 . . . . . . . . . . . . . . . . . . . . . . . . . . . 2-6

2.2.3 Live Load Reduction for Columns, Beams, and Slabs . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 2-7
2.2.3.1 Example: Live Load Reductions for Building #2 . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 2-8
2.2.4 Factored Loads . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 2-9
2.3 FRAME ANALYSIS BY COEFFICIENTS . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 2-11
2.3.1 Continuous Beams and One-Way Slabs. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 2-11
2.3.2 Example: Frame Analysis by Coefficients . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 2-13
2.4 FRAME ANALYSIS BY ANALYTICAL METHODS . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 2-14
2.4.1 Stiffness . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 2-14
2.4.2 Arrangement of Live Load. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 2-15
2.4.3 Design Moments . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 2-15
2.4.4 Two-Cycle Moment Distribution Analysis for Gravity Loading . . . . . . . . . . . . . . . . . . . . . . . 2-18
2.5 COLUMNS . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 2-18
2.6 LATERAL LOAD ANALYSIS. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 2-19
2.6.1 Portal Method . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 2-19
2.6.2 Examples: Wind Load Analyses for Buildings #1 and #2 . . . . . . . . . . . . . . . . . . . . . . . . . . . 2-20
References . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 2-26
Chapter 3—Simplified Design for Beams and One-Way Slabs . . . . . . . . . . . . . . . . . . . . . 3-1
3.1 INTRODUCTION . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 3.1
3.2 DEPTH SELECTION FOR CONTROL OF DEFLECTIONS . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 3-1
3.3 MEMBER SIZING FOR MOMENT STRENGTH . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 3-3
3.3.1 Notes on Member Sizing for Economy . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 3-4
3.4 DESIGN FOR FLEXURAL REINFORCEMENT . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 3-5
3.5 REINFORCING BAR DETAILS . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 3-6
3.6 DESIGN FOR SHEAR REINFORCEMENT . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 3-11
3.6.1 Example: Design for Shear Reinforcement . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 3-20
3.6.2 Selection of Stirrups for Economy . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 3-22
3.7 DESIGN FOR TORSION . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 3-22
3.7.1 Beam Sizing to Neglect Torsion . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 3-23
3.7.1.1 Example: Beam Sizing to Neglect Torsion . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 3-30
3.7.2 Beam Design Considering Torsion . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 3-31
3.7.3 Simplified Design for Torsion Reinforcement . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 3-32
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3.7.3.1 Example: Design for Torsion Reinforcement . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 3-34
3.8 EXAMPLES: SIMPLIFIED DESIGN FOR BEAMS AND ONE-WAY SLABS. . . . . . . . . . . . . . . . . 3-36
3.8.1 Example: Design of Standard Pan Joists for Alternate (1) Floor System
(Building #1) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 3-37
3.8.2 Example: Design of Wide-Module Joists for Alternate (2) Floor System (Building #1) . . . . 3-41
3.8.3 Example: Design of the Support Beams for the Standard Pan Joist Floor Along a
Typical N-S Interior Column Line (Building #1) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 3-46
References . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 3-52
Chapter 4—Simplified Design for Two-Way Slabs. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 4-1
4.1 INTRODUCTION . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 4-1
4.2 DEFLECTION CONTROL–MINIMUM SLAB THICKNESS . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 4-3
4.3 TWO-WAY SLAB ANALYSIS BY COEFFICIENTS—DIRECT DESIGN METHOD . . . . . . . . . . . . . 4-6
4.4 SHEAR IN TWO-WAY SLAB SYSTEMS . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 4-15
4.4.1 Shear in Flat Plate and Flat Slab Floor Systems . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 4-16
4.5 COLUMN MOMENTS DUE TO GRAVITY LOADS . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 4-21
4.6 REINFORCEMENT DETAILING . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 4-23
4.7 EXAMPLES: SIMPLIFIED DESIGN FOR TWO-WAY SLABS . . . . . . . . . . . . . . . . . . . . . . . . . . . 4-23
4.7.1 Example: Interior Strip (N-S Direction) of Building #2, Alternate (2) . . . . . . . . . . . . . . . . . . 4-24
4.7.2 Example: Interior Strip (N-S Direction) of Building #2, Alternate (1) . . . . . . . . . . . . . . . . . . 4-31
References . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 4-45
Chapter 5—Simplified Design for Columns . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 5-1
5.1 INTRODUCTION . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 5-1
5.2 DESIGN CONSIDERATIONS. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 5-1
5.2.1 Column Size . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 5-1
5.2.2 Column Constructability . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 5-1
5.2.3 Column Economics . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 5-2
5.3 DESIGN STRENGTH FOR COLUMNS . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 5-2
5.4 PRELIMINARY COLUMN SIZING . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 5-4
5.5 SIMPLIFIED DESIGN FOR COLUMNS. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 5-7
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5.5.1 Simplified Design Charts—Combined Axial Load and Bending Moment . . . . . . . . . . . . . . . . 5-7
5.5.1.1 Example: Construction of Simplified Design Chart. . . . . . . . . . . . . . . . . . . . . . . . . . . 5-11
5.5.2 Column Ties . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 5-15
5.5.3 Biaxial Bending of Columns . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 5-17
5.5.3.1 Example: Simplified Design of a Column Subjected to Biaxial Loading . . . . . . . . . . 5-18
5.6 COLUMN SLENDERNESS CONSIDERATIONS. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 5-22
5.6.1 Non-sway versus Sway Frames . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 5-22
5.6.2 Minimum Sizing for Design Simplicity. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 5-22
5.7 PROCEDURE FOR SIMPLIFIED COLUMN DESIGN . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 5-24
5.8 EXAMPLES: SIMPLIFIED DESIGN FOR COLUMNS . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 5-25
5.8.1 Example: Design of an Interior Column Stack for Building #2 Alternate (1)—Slab and
Column Framing Without Structural Walls (Sway Frame). . . . . . . . . . . . . . . . . . . . . . . . . . . 5-25
5.8.2 Example: Design of an Interior Column Stack for Building #2 Alternate (2)—Slab and
Column Framing with Structural Walls (Non-sway Frame) . . . . . . . . . . . . . . . . . . . . . . . . . . 5-29
5.8.3 Example: Design of an Edge Column Stack (E-W Column Line) for Building #1—
3-story Pan Joist Construction (Sway Frame). . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 5-31
5.9 COLUMN SHEAR STRENGTH . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 5-35
5.9.1 Example: Design for Column Shear Strength . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 5-35
References . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 5-45
Chapter 6—Simplified Design for Structural Walls . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 6-1
6.1 INTRODUCTION . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 6-1
6.2 FRAME-WALL INTERACTION. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 6-1
6.3 WALL SIZING FOR LATERAL BRACING . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 6-2
6.3.1 Example: Wall Sizing for Non-sway Condition. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 6-2
6.4 DESIGN FOR SHEAR . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 6-4
6.4.1 Example 1: Design for Shear. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 6-7
6.4.2 Example 2: Design for Shear. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 6-9
6.5 DESIGN FOR FLEXURE . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 6-11
6.5.1 Example: Design for Flexure . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 6-12
References . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 6-20
x

Chapter 7—Simplified Design for Footings . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 7-1
7.1 INTRODUCTION . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 7-1
7.2 PLAIN CONCRETE VERSUS REINFORCED CONCRETE FOOTINGS . . . . . . . . . . . . . . . . . . . . 7-1
7.3 SOIL PRESSURE. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 7-1
7.4 SURCHARGE. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 7-2
7.5 ONE-STEP THICKNESS DESIGN FOR REINFORCED CONCRETE FOOTINGS . . . . . . . . . . . . 7-3
7.5.1 Procedure for Simplified Footing Design . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 7-5
7.6 FOOTING DOWELS. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 7-12
7.6.1 Vertical Force Transfer at Base of Column . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 7-12
7.6.2 Horizontal Force Transfer at Base of Column . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 7-16
7.7 EXAMPLE: REINFORCED CONCRETE FOOTING DESIGN . . . . . . . . . . . . . . . . . . . . . . . . . . . 7-16
7.8 ONE-STEP THICKNESS DESIGN FOR PLAIN CONCRETE FOOTINGS . . . . . . . . . . . . . . . . . . 7-20
7.8.1 Example: Plain Concrete Footing Design . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 7-21
References . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 7-22
Chapter 8—Structural Detailing of Reinforcement for Economy . . . . . . . . . . . . . . . . . . . 8-1
8.1 INTRODUCTION . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 8-1
8.2 DESIGN CONSIDERATIONS FOR REINFORCEMENT ECONOMY . . . . . . . . . . . . . . . . . . . . . . . 8-1
8.3 REINFORCING BARS . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 8-3
8.3.1 Coated Reinforcing Bars. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 8-3
8.4 DEVELOPMENT OF REINFORCING BARS . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 8-4
8.4.1 Introduction . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 8-4
8.4.2 Development of Deformed Bars in Tension . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 8-4
8.4.3 Development of Hooked Bars in Tension . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 8-7
8.4.4 Development of Bars in Compression . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 8-8
8.5 SPLICES OF REINFORCING BARS . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 8-8
8.5.1 Tension Lap Splices . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 8-9
8.5.2 Compression Lap Splices . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 8-9
8.6 DEVELOPMENT OF FLEXURAL REINFORCEMENT. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 8-11
8.6.1 Introduction . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 8-11
xi

8.6.2 Requirements for Structural Integrity . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 8-11
8.6.3 Recommended Bar Details . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 8-11
8.7 SPECIAL BAR DETAILS AT SLAB-TO-COLUMN CONNECTIONS . . . . . . . . . . . . . . . . . . . . . . . 8-12
8.8 SPECIAL SPLICE REQUIREMENTS FOR COLUMNS. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 8-16
8.8.1 Construction and Placing Considerations. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 8-16
8.8.2 Design Considerations . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 8-16
8.8.3 Example: Lap Splice Length for an Interior Column of Building #2, Alternate (2) Slab and
Column Framing with Structural Walls (Non-sway Frame) . . . . . . . . . . . . . . . . . . . . . . . . . . 8-17
Column Framing Without Structural Walls (Sway Frame) . . . . . . . . . . . . . . . . . . . . . . . . . . .8-20
References . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 8-21
Chapter 9—Design Considerations for Economical Formwork . . . . . . . . . . . . . . . . . 9-1
9.1 INTRODUCTION . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 9-1
9.2 BASIC PRINCIPLES TO ACHIEVE ECONOMICAL FORMWORK . . . . . . . . . . . . . . . . . . . . . . . . . 9-1
9.2.1 Standard Forms . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 9-1
9.2.2 Repetition . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 9-2
9.2.3 Simplicity. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 9-2
9.3 ECONOMICAL ASPECTS OF HORIZONTAL FRAMING . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 9-3
9.3.1 Slab Systems . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 9-3
9.3.2 Joist Systems . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 9-5
9.3.3 Beam-Supported Slab Systems . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 9-5
9.4 ECONOMICAL ASPECTS OF VERTICAL FRAMING . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 9-6
9.4.1 Walls . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 9-6
9.4.2 Core Areas . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 9-6
9.4.3 Columns . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 9-6
9.5 GUIDELINES FOR MEMBER SIZING . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 9-6
9.5.1 Beams. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 9-6
9.5.2 Columns . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 9-8
9.5.3 Walls . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 9-9
9.6 OVERALL STRUCTURAL ECONOMY . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 9-9
References . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 9-10
xii

Chapter 10—Design Considerations for Fire Resistance. . . . . . . . . . . . . . . . . . . . . . . . . . 10-1
10.1 INTRODUCTION . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 10-1
10.2 DEFINITIONS. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 10-2
10.3 FIRE RESISTANCE RATINGS. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 10-2
10.3.1 Fire Test Standards. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 10-2
10.3.2 ASTM E 119 Test Procedure. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 10-3
10.4 DESIGN CONSIDERATIONS FOR FIRE RESISTANCE. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 10-4
10.4.1 Properties of Concrete . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 10-4
10.4.2 Thickness Requirements. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 10-5
10.4.3 Cover Requirements . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 10-5
10.5 MULTICOURSE FLOORS AND ROOFS . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 10-8
10.5.1 Two-Course Concrete Floors . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 10-8
10.5.2 Two-Course Concrete Roofs. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 10-9
10.5.3 Concrete Roofs with Other Insulating Materials . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 10-9
Reference . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 10-10
Chapter 11— Design Considerations for Earthquake Forces . . . . . . . . . . . . . . . . . . . . . . 11-1
11.1 INTRODUCTION. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 11-1
11.2 SEISMIC DESIGN CATEGORY (SDC) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 11-1
11.3 REINFORCED CONCRETE EARTHQUAKE-RESISTING STRUCTURAL SYSTEMS . . . . . . . . 11-6
11.4 STRUCTURES EXEMPT FORM SEISMIC DESIGN REQUIREMENTS . . . . . . . . . . . . . . . . . . . 11-7
11.5 EARTHQUAKE FORCES . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 11-8
11.6 EQUIVALENT LATERAL FORCE PROCEDURE . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 11-9
11.6.1 Design Base Shear . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 11-9
11.6.2 Vertical Distribution of Seismic Forces . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 11-10
11.6.2.1 Distribution of Seismic Forces to Vertical Elements of the Lateral Force
Resisting System . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 11-12
11.6.2.2 Direction of Seismic Load . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 11-14
11.6.3 Load Combinations for Seismic Design . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 11-15
11.7 OVERTURNING . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 11-15
11.8 STORY DRIFT . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 11-15
xiii

11.9 P-⌬ EFFECT . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 11-16
11.10 DESIGN AND DETAILING REQUIREMENTS . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 11-17
11.11 EXAMPLES . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 11-18
11.11.1 Example 1 – Building # 2 Alternate (2) Shearwalls . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 11-18
11.11.2 Example 2 – Building # 1 Alternate (1) Standard Pan Joist . . . . . . . . . . . . . . . . . . . . . . 11-31
References. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 11-38
Chapter 12— Introduction to Sustainable Design . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 12-1
xiv

Direct all correspondence to:
Buildings
Portland Cement Association
5420 Old Orchard Road
Skokie, Illinois 60077-1083
Tel: 847.966.6200
Fax: 847.966.9781
Email: info@cement.org

About the authors …
Mahmoud Kamara, PhD is PCA’s senior structural engineer for engineered buildings.
He is the author and coauthor of numerous PCA publications and technical guides. Kamara
serves on several technical structural committees and is an active member of the American
Concrete Institute and the American Society of Civil Engineers and he chairs ACI/ASCE
Joint Committee 421 Design of Reinforced Concrete Slabs. Prior to joining PCA,
Kamara held faculty positions at the University of Alexandria, Egypt and the University of
Alabama at Birmingham. His experience includes teaching structural engineering courses at
the undergraduate and graduate levels, performing research in the areas of reinforced concrete
and concrete technology, structural design and consulting, software developing and structural
forensic investigations. He received the ACI Structural Research Award in 1992 and is
the recipient of the Structural Engineers Association of Illinois’ Meritorious Publication
Award for 2008.
Lawrence C. Novak, SE, FACI, LEED® AP, has more than 25 years of experience as a
structural engineer on high-rise, mid-rise and special use structures throughout the world,
including seismic regions. He is the Director of Engineered Buildings with the Portland
Cement Association. Prior to joining the PCA, he was an Associate Partner with Skidmore,
Owings & Merrill where he recently served as the lead structural engineer for the Burj Dubai
Tower, the world’s tallest building. Novak serves on several technical structural committees
and is an active member of the American Concrete Institute including the ACI 318 on the
Structural Concrete Building Code, ACI 445 on Shear and Torsion, ACI 445-A on Strut and
Tie Modeling, ACI 209 on Creep and Shrinkage and ACI 130 on Sustainability of Concrete.
He has served on the Board of Directors of several engineering organizations including
SEAOI, TCA and the Illinois Engineering Hall of Fame. He has co-authored numerous
publications on structural engineering and is the recipient of the ASCE Illinois Chapter
“Citizen Engineer of the Year” award, Structural Engineers Association of Illinois’
“Meritorious Publication Award” for 2001, 2008 and 2009, the National Council of Structural
Engineers Associations’ “Outstanding Structural Engineering Publication” Award for 2001
and the United Kingdom’s Oscar Faber Award for 2002. In addition to being a Licensed
Structural Engineer, Mr. Novak is a LEED® Accredited Professional and a Certified Structural
Peer Reviewer.

1-1
Chapter 1
A Simplified Design Approach
1.1 THE BUILDING UNIVERSE
There is a little doubt that the construction of a very tall building, a large domed arena, or any other prominent
mega structure attracts the interest of a great number of structural engineers. The construction of such
structures usually represents the highest level of sophistication in structural design and often introduces
daring new concepts and structural innovations as well as improvements in construction techniques.
Many structural engineers have the desire to become professionally involved in the design of such distinctive
buildings during their careers. However, very few projects of this prestigious caliber are built in any given year.
Truly, the building universe consists largely of low-rise and small-area buildings. Figure 1-1 shows the
percentage of building floor area constructed in 2002 in terms of different building height categories. The
figure shows that the vast majority of the physical volume of construction is in the 1- to 3-story height range.
1 to 3 floors (93%) 1 to 3 floors (88%)
All Buildings Nonresidential Buildings
4 to 15 floors (6%)
4 to 15 floors (9%)
> 15 floors (3%)> 15 floors (1%)
Figure 1-1 Floor Area of Construction, 2002

Simplified Design • EB204
1-2
In the same way, Figure 1-2 shows the percentage of all building projects constructed in various size
categories. Building projects less than 15,000 sq ft dominate the building market.
When all these statistics are considered, it becomes quickly apparent that while most engineers would like to
work on prestigious and challenging high-rise buildings or other distinctive structures, it is far more likely that
they will be called upon to design smaller and shorter buildings.
1.2 COST EFFICIENCIES
The benefit of efficient materials use is not sought nor realized in a low-rise building to the same degree as in a
high-rise building. For instance, reducing a floor system thickness by an inch may save three feet of building height
in a 36-story building and only 3 in. in a three-story building. The added design costs needed to make thorough
studies in order to save the inch of floor depth may be justified by construction savings in the case of the 36-story
building, but is not likely to be warranted in the design of the shorter building. As a matter of fact, the use of more
materials in the case of the low-rise building may sometimes enable the engineer to simplify construction features and
thereby effectively reduce the overall cost and time required to complete the building.
In reviewing cost studies of several nonresidential buildings, it was also noted that the cost of a building’s frame and
envelope represent a smaller percentage of the total building cost in low-rise buildings than in high-rise buildings.
In low-rise concrete construction, designs that seek to simplify concrete formwork will probably result in more
economical construction than those that seek to optimize the use of reinforcing steel and concrete, since
forming represents a significant part of the total frame costs. There is less opportunity to benefit from form
repetition in a low-rise building than in a high-rise building.
< 15,000 sq ft(98%)
> 50,000 sq ft(1%)
15,000 to 50,000 sq ft(1%)
Figure 1-2 All Building Project Size. 2002

Considering the responsibility of the engineer to provide a safe and cost-effective solution to the needs of the
building occupant and owner, it becomes clear that, for the vast majority of buildings designed each year, there
should be an extra effort made to provide for expediency of construction rather than efficiency of structural
design. Often, the extra time needed to prepare the most efficient designs with respect to structural materials
is not justified by building cost or performance improvements for low-rise buildings.
1.3 THE COMPLEX CODE
In 1956 the ACI 318 Code was printed in 73 small-size pages; by 2011, ACI 318 and Commentary contained
more than 500 large-size pages of Code and Commentary—a very substantial increase in the amount of
printed material with which an engineer has to become familiar in order to design a concrete building.
Furthermore, the code revision in 2002 has seen the most significant technical revisions since 1963. Several
new and sweeping changes were introduced in the load and resistance factors, unified design provisions for
reinforced and prestressed concrete design were introduced. For the first time, a new appendix on anchorage
to concrete is provided along with another appendix on the strut-and-tie modeling and design techniques.
To find the reasons for the proliferation in code design requirements since 1956, it is useful to examine the extensive
changes in the makeup of some of the buildings that required and prompted more complex code provisions.
1.3.1 Complex Structures Require Complex Designs
Advances in the technology of structural materials, new building systems, and new engineering procedures
have resulted in the use of concrete in a new generation of more flexible structures, dramatically different from
those for which the old codes were developed.
In the fifties, 3000 psi concrete was the standard in the construction industry. Today, concrete with 12,000 psi
to 16,000 psi and higher strength is used for lower story columns and walls of very tall high-rise
buildings. Grade 40 reinforcing steel has almost entirely been replaced by Grade 60 reinforcement. Today the
elastic modulus of concrete plays an equally important role to compressive strength in building deflection
calculations and serviceability checks.
Gradual switching in the 1963 and 1971 Codes from the Working Stress Design Method to the Strength Design
Method permitted more efficient designs of the structural components of buildings. The size of structural
sections (columns, beams and slabs) became substantially smaller and utilized less reinforcement, resulting in
a 20 to 25% reduction in structural frame costs. In 2002 the working stress design method, long in Appendix A as
an alternate design method was deleted.
While we have seen dramatic increases in strength of materials and greater cost efficiencies and design inno-
vations made possible by the use of strength design method, we have, as a consequence, also created new and
more complex problems. The downsizing of structural components has reduced overall building stiffness.
A further reduction has resulted from the replacement of heavy exterior cladding and interior partitions with
1-3
Chapter 1 • A Simplified Design Approach

1-4
lightweight substitutes, which generally do not contribute significantly to building mass or stiffness.
In particular, the drastic increase of stresses in the reinforcement at service loads from less than 20 ksi to
nearly 40 ksi has caused a significantly wider spread of flexural cracking at service loads in slabs and beams,
with consequent increases in their deflections.
When structures were designed by the classical working stress approach, both strength and serviceability
of the structure were ensured by limiting the stresses in the concrete and the reinforcement, in addition to
imposing limits on slenderness ratios of the members. The introduction of strength design with the resulting
increase in member slenderness significantly lengthened the design process; in addition to designing for
strength, a separate consideration of serviceability (deflections and cracking) became necessary.
We are now frequently dealing with bolder, larger, taller structures, which are not only more complex, but also
more flexible. They are frequently for mixed use and as a result comprise more than one building and floor
system. Their structural behavior is characterized by larger deformations relative to member dimensions than
we had experienced in the past. As a consequence, a number of effects which heretofore were considered
secondary and could be neglected now become primary considerations during the design process. In this
category are changes in geometry of structures due to gravity and lateral loadings. The effects of shrinkage,
creep, and temperature are also becoming significant and can no longer be neglected in tall or long structures
because of their cumulative effects. Seismic codes continue to evolve and consolidate demanding consistent
risk assessment and demanding more aggressive design and detailing requirements.
Building and material codes are consensus documents written and edited by committees which can lead to
complications. Such committee is often hampered by the legal language these codes need in order to be adopted
as a law. This format restricts what can be said and how to say it, which results in a complicated document that
is not intended for easy reading and understanding.
1.4 A SIMPLE CODE
The more complex buildings undoubtedly require more complex design procedures to produce safe and
economical structures. However, when we look at the reality of the construction industry as discussed at the
beginning of this chapter, it makes little sense to impose on structures of moderate size and height intricate
design approaches that were developed to assure safety in high complex structures. While the advances of the
past decades have made it possible to build economical concrete structures soaring well over quarter mile in
height, the makeup of low-rise buildings has not changed all that significantly over the years.
It is possible to write a simplified code to be applicable to both moderately sized structures and large complex
structures. However, this would require a technical conservatism in proportioning of members. While the cost
of moderate structures would not be substantially affected by such an approach, the competitiveness of large
complex structures could be severely impaired. To avoid such unnecessary penalties, and at the same time to
stay within required safety limits, it is possible to extract from the complex code a simplified design approach
that can be applied to specifically defined moderately sized structures. Such structures are characterized as
having configurations and rigidity that eliminate sensitivity to secondary stresses and as having members
proportioned with sufficient conservatism to be able to simplify complex code provisions.

1.5 PURPOSE OF SIMPLIFIED DESIGN
A complex code is unavoidable since it is necessary to address simple and complex structures in the same
document. The purpose of this book is to give practicing engineers some way of reducing the design time
required for smaller projects, while still complying with the letter and intent of the ACI Standard 318, Building
Code Requirements for Structural Concrete.1.1 The simplification of design with its attendant savings in design
time result from avoiding building member proportioning details and material property selections which make
it necessary to consider certain complicated provisions of the ACI Standard. These situations can often
be avoided by making minor changes in the design approach. In the various chapters of this book, specific
recommendations are made to accomplish this goal.
The simplified design procedures presented in this manual are an attempt to satisfy the various design
considerations that need to be addressed in the structural design and detailing of primary framing members of
a reinforced concrete building—by the simplest and quickest procedures possible. The simplified design
material is intended for use by experienced engineers well versed in the design principles of reinforced
concrete and completely familiar with the design provisions of ACI 318. It aims to arrange the information in
the code in an organized step-by-step procedure for the building and member design. The formulae and
language avoid complicated legal terminology without changing the intent or the objective of the code.
As noted above, this book is written solely as a time saving design aid; that is, to simplify design procedures
using the provisions of ACI 318 for reinforced concrete buildings of moderate size and height.
1.6 SCOPE OF SIMPLIFIED DESIGN
The simplified design approach presented in this book should be used within the general guidelines and
limitations given in this section. In addition, appropriate guidelines and limitations are given within each
chapter for proper application of specific simplifying design procedures.
• Type of Construction: Conventionally reinforced cast-in-place construction. Prestressed and precast
construction are not addressed.
• Building Size: Buildings of moderate size and height with usual spans and story heights. Maximum building plan
dimension should not exceed 250 ft to reduce effects of shrinkage and temperature to manageable levels.1.2
Maximum
building height should not exceed 7 stories to justify the economics of simplified design.
• Building Geometry: Buildings with standard shapes and geometry suitable for residential and commercial
occupancies. Buildings should be free of plan and vertical irregularities as defined in most building codes
(see chapter 11). Such irregularities demand more rigorous building analysis and detailed member design.
1-5

1-6
• Materials: Normal weight concrete › = 4000 psi
Deformed reinforcing bars fy = 60,000 psi
Both material strengths are readily available in the market place and will result in members that are durable* and
perform structurally well. Traditionally cost analyses have shown that for gravity loads, conventionally reinforced
concrete floor systems with › = 4000 psi are more economical than ones with higher concrete strengths.1.3
One set of material parameters greatly simplifies the presentation of design aids. The 4000/60,000 strength
combination is used in all simplified design expressions and design aids presented in this book with the following
exceptions: the simplified thickness design for footings and the tables for development lengths consider concrete
compressive strength of › = 3000 psi and › = 4000 psi. The use of › = 4000 psi corresponds to a standard β1 = 0.85.
Further simplification is achieved in the minimum reinforcement equations by excluding the term .
In most cases, the designer can easily modify the simplified design expressions for other material
strengths. Throughout the book coefficients, tables, and graphs covering other concrete strengths are given
to help designers expand the scope of the design concept presented in the publication. Also, welded wire
reinforcement and lightweight concrete may be used with the simplified design procedures, with appropriate
modification as required by ACI 318.
• Loading: Design dead load, live load, seismic and wind forces are in accordance with American Society of Civil
Engineers Minimum Design Loads for Buildings and Other Structures (ASCE/SEI 7-05)1.5
, with reductions in
live loads as permitted in ASCE/SEI 7-05. The building code having jurisdiction in the locality of construction
should be consulted for any possible differences in design loads from those given in ASCE/SEI 7-05.
If resistance to earth or liquid pressure, impact effects, or structural effects of differential settlement,
shrinkage, or temperature change need to be included in design, such effects are to be included separately,
in addition to the effects of dead load, live load, and lateral forces (see ACI 9.2). Also, effects of forces
due to snow loads, rain loads (ponding), and fixed service equipment (concentrated loads) are to be
considered separately where applicable (ACI 8.2). Exposed exterior columns or open structures may require
consideration of temperature change effects, which are beyond the scope of this manual. Additionally, the
durability requirements given in ACI Chapter 4 must be considered in all cases (see Section 1.7 of this book).
• Design Method: All simplified design procedures comply with provisions of Building Code Requirements
for Structural Concrete (ACI 318-11), using appropriate load factors and strength reduction factors as
specified in ACI 9.2 and 9.3. References to specific ACI Code provisions are noted (e.g., ACI 9.2 refers to
ACI 318-11, Section 9.2).
* This applies to members which are not exposed to 1) freezing and thawing in a moist condition, 2) deicing chemicals and 3) severe
levels of sulfates (see ACI Chapter 4).
ʹfc

1-7
1.7 BUILDING EXAMPLES
To illustrate application of the simplified design approach presented in this book, two simple and regular
shaped building examples are included. Example No. 1 is a 3-story building with one-way joist slab and
column framing. Two alternate joist floor systems are considered: (1) standard pan joist and (2) wide-module
pan joist. The beam column frame is used for lateral force resistance. Example No. 2 is a 5-story building with
two-way flat plate and column framing. Two alternate lateral-force resisting systems are considered: (1) slab
and column framing with spandrel beams and (2) structural shearwalls. In all cases, it is assumed that the
members will not be exposed to freezing and thawing, deicing chemicals and severe levels of sulfates since
this is the case in most enclosed multistory buildings. Therefore, a concrete compressive strength of › = 4000
psi can be used for all members. ACI Chapter 4 should be consulted if one or more of these aspects must be
considered. In some cases, › must be larger than 4000 psi to achieve the required durability.
To illustrate simplified design, typical structural members of the two buildings (beams, slabs, columns, walls,
and footings) are designed by the simplified procedures presented in the various chapters of this book.
Guidelines for determining preliminary member sizes and required fire resistance are given in Section 1.8.

1.7.1 Building No. 1—3-Story Pan Joist Construction
(1) Floor system: one-way joist slab
Alternate (1)—standard pan joists
Alternate (2)—wide-module joists
(2) Lateral-force resisting system: beam and column framing
(3) Load data: roof LL = 12 psf
DL = 105 psf (assume 95 psf joists and beams + 10 psf roofing and misc.)
floors LL = 60 psf
DL = 130 psf (assume 100 psf joists and beams + 20 psf partitions +
10 psf ceiling and misc.)
(4) Preliminary sizing:
Columns interior = 18 ϫ 18 in.
exterior = 16 ϫ 16 in.
Width of spandrel beams = 20 in.
Width of interior beams = 36 in.
(5) Fire resistance requirements:
floors: Alternate (1)—1 hour
Alternate (2)—2 hours*
roof: 1 hour
columns: 1 hour**
Figure 1-3 shows the plan and elevation of Building #1.
* In some cases, floors may be serving as an “occupancy separation” and may require a higher rating based on building type of
construction. For example, there may be a mercantile or parking garage on the lowest floor.
** Columns and walls supporting two hour rated floor, as in Alternate (2), are required to have a two hour rating.
1-8

3@13'-0"=39'-0"
N
5 @ 30'-0" = 150'-0"
3@30'-0"=90'-0"
Wide-Module
Joists–
Alternate (2)
Standard Pan
Joists–
Alternate (1)
Figure 1-3 Plan and Elevation of Building #1
1-9

* Assume 20 psf partitions + 10 psf ceiling and misc.
** Assume interior portions of walls enclose exit stairs.
1-10
1.7.2 Building No. 2—5-Story Flat Plate Construction
(1) Floor system: two-way flat plate – with spandrel beams for Alternate (1)
(2) Lateral-force resisting system
Alternate (1)—slab and column framing-with spandrel beam
Alternate (2)—structural shearwalls
(3) Load data: roof LL = 20 psf
DL = 122 psf
floors LL = 50 psf
DL* = 142 psf (9 in. slab)
136 psf (8.5 in. slab)
(4) Preliminary sizing:
Slab (with spandrels) = 8.5 in.
Slab (without spandrels) = 9 in.
Spandrels = 12 ϫ 20 in.
(5) Fire resistance requirements:
floors: 2 hours
roof: 1 hour
columns: 2 hours
shearwalls:** 2 hours
Figure 1-4 shows the plan and elevation of Building #2.

15'-0"4@12'-0"=48'-0"
Exit stair (typ.)
Shearwalls
Alternate (2)
N
Spandrel Beams
Alternate (1)
5 @ 24'-0" = 120'-0"
3@20'-0"=60'-0"
Figure 1-4 Plan and Elevation of Building #2
1-11

1-12
1.8 PRELIMINARY DESIGN
Preliminary member sizes are usually required to perform the initial frame analysis and/or to obtain initial
quantities of concrete and reinforcing steel for cost estimating. Practical initial member sizes are necessary
even when a computer analysis is used to determine the load effects on a structure. The guidelines for
preliminary design given in the following sections are applicable to regular buildings of moderate size and
height. These guidelines were used to obtain the preliminary sizes listed in Sections 1.7.1 and 1.7.2 for the two
example buildings. Chapters 8 and 9 list additional guidelines to achieve overall economy.
1.8.1 Floor Systems
Various factors must be considered when choosing a floor system. The magnitude of the superimposed loads
and the bay size (largest span length) are usually the most important variables to consider in the selection
process. Fire resistance is also very important (see Section 1.8.5). Before specifying the final choice for the
floor system, it is important to ensure that it has at least the minimum fire resistance rating prescribed in
the governing building code.
In general, different floor systems have different economical span length ranges for a given total factored load.
Also, each system has inherent advantages and disadvantages, which must be considered for a particular
project. Since the floor system (including its forming) accounts for a major portion of the overall cost of a
structure, the type of system to be utilized must be judiciously chosen in every situation.
Figures 1-5 through 1-7 can be used as a guide in selecting a preliminary floor system with › = 4000 psi.1.3
A relative cost index and an economical square bay size range are presented for each of the floor systems listed.
In general, an exact cost comparison should be performed to determine the most economical system for a
given building.
Once a particular floor system has been chosen, preliminary sizes must be determined for the members in the
system. For one-way joists and beams, deflection will usually govern. Therefore, ACI Table 9.5(a) should be
used to obtain the preliminary depth of members that are not supporting or attached to partitions and other
construction likely to be damaged by deflection. The width of the member can then be determined by the
appropriate simplified equation given in Chapter 3. Whenever possible, available standard sizes should
be specified; this size should be repeated throughout the entire structure as often as possible. For overall
economy in a standard joist system, the joists and the supporting beams must have the same depth. This also
provides an optimum ceiling cavity to a uniform bottom of floor elevation with maximum clearance for building
mechanical/electrical/plumbing (M/E/P) systems.
For flat plate floor systems, the thickness of the slab will almost always be governed by two-way (punching)
shear. Figures 1-8 through 1-10 can be used to obtain a preliminary slab thickness based on two-way shear
at an interior square column and › = 4000 psi. For a total factored load qu (psf) and the ratio of the floor
tributary area, A, to the column area c1
2, a value of d/c1, can be obtained from the figure. Note that d is the dis-
tance from the compression face of the slab to the centroid of the reinforcing steel. The preliminary thickness
of the slab h can be increased by adding 1.25 in. to the value of d (see Chapter 4).
It is important to note that the magnitude of the unbalanced moment at an interior column is usually small.
However, at an edge column, the shear stress produced by the unbalanced moment can be as large as or larger than
the shear stress produced by the direct shear forces. Consequently, refined calculations to account for the effect
of the unbalanced moment should be done according to Chapter 4.

1-13
CostIndex
Square Bay Size (ft)
20 25 30 35 40 45 50
0.65
1.40
1.35
1.30
1.25
1.20
1.15
1.10
1.05
1.00
0.95
0.90
0.85
0.80
0.75
0.70
Flat Plate
Flat Slab
One-Way Joist (30" pan)
Two-Way Joist (3' module)
Figure 1-5 Relative Costs of Reinforced Concrete Floor Systems, Live Load = 50 psf1.3

1-14
CostIndex
20 25 30 35 40 45 50
0.65
1.40
1.35
1.30
1.25
1.20
1.15
1.10
1.05
1.00
0.95
0.90
0.85
0.80
0.75
0.70
Flat Plate
Flat Slab
Figure 1-6 Relative Costs of Reinforced Concrete Floor Systems, Live Load = 100 psf1.3

1-15
Flat Plate100 Flat Slab One-Way Joist
Flat Plate Flat Slab One-Way Joist50
LiveLoad(psf)
20 25 30 35 40 45 50
Figure 1-7 Cost-Effective Concrete Floor Systems as a Function of Span Length and Load1.3

1-16
When increasing the overall slab thickness is not possible or feasible, drop panels or shear caps can be provided at
the column locations where two-way shear is critical. Chapter 4 gives ACI 318 provisions for minimum drop
panel dimensions.
In all cases, the slab thickness must be larger than the applicable minimum thickness given in ACI 9.5.3 for
deflection control. Figure 4-4 may be used to determine the minimum thickness as a function of the clear span
n
for the various two-way systems shown.
1.8.2 Columns
For overall economy, the dimensions of a column should be determined for the load effects in the lowest story of
the structure and should remain constant for the entire height of the building; only the amounts of reinforcement
should vary with respect to height.* The most economical columns usually have reinforcement ratios in the range
of 1-2%. In general, it is more efficient to increase the column size than to increase the amount of reinforcement.
This approach is taken to eliminate congestion of column reinforcement, which has to be critically evaluated along
the lap splice length and to accommodate horizontal beam bars at the beam column intersections.
Columns in a frame that is braced by shearwalls (non-sway frame) are subjected to gravity loads only. Initial
column sizes may be obtained from design aids such as the one given in Fig. 5-2: assuming a reinforcement ratio
in the range of 1-2%, a square column size can be determined for the total factored axial load Pu in the lowest
story. Once an initial size is obtained, it should be determined if the effects of slenderness need to be considered.
If feasible, the size of the column should be increased so as to be able to neglect slenderness effects.
When a frame is not braced by shearwalls (sway frame), the columns must be designed for the combined effects
of gravity and wind loads. In this case, a preliminary size can be obtained for a column in the lowest level from
Fig. 5-2 assuming that the column carries gravity loads only. The size can be chosen based on 1% reinforcement
in the column; in this way, when wind loads are considered, the area of steel can usually be increased without
having to change the column size. The design charts given in Figs. 5-18 through 5-25 may also be used to
determine the required column size and reinforcement of a given combination of factored axial loads and
moments. Note that slenderness effects can have a significant influence on the amount of reinforcement that is
required for a column in a sway frame; for this reason, the overall column size should be increased (if possible)
to minimize the effects of slenderness.
1.8.3 Shearwalls
For buildings of moderate size and height, a practical range for the thickness of shearwalls is 8 to 10 in.
The required thickness will depend on the length of the wall, the height of the building, and the tributary wind
area of the wall. In most cases, minimum amounts of vertical and horizontal reinforcement are sufficient for
both shear and moment resistance.
In the preliminary design stage, the shearwalls should be symmetrically located in the plan (if possible) so that
torsional effects on the structure due to lateral loads are minimized.
* In tall buildings, › may vary along the height as well.

0.10
0.15
0.20
0.25
0.30
0.35
0.40
0.45
0.50
0.55
0.60
0.65
0.70
0.75
0.80
0.85
200 220 240 260 280 300 320 340 360 380 400
Wu, psf
100
125
150
175
200
225
250
275
300
325
350
375
A = 400
c2
1
˜2
˜1c1
c1
A = ˜1 x ˜2
Panel
centerline
qu, psf
d/c1
Figure 1-8 Preliminary Design Chart for Slab Thickness Based on Two-Way Shear at an Interior
Square Column (› = 4000 psi)—Unbalanced moment transfer is not considered
1-17

1-18
1
0.9
0.8
0.7
0.6
0.5
0.4
0.3
0.2
0.1
200 220 240 260 280 300 320 340 360 380 400
qu, psf
d/c1
c1
c1
˜2
˜1/2 + C1/2
A= ˜2(˜1/2 + C1/2)
Panel
centerline
350
375
300
275
250
225
200
175
150
125
100
75
50
A
c2
1
=
Figure 1-9 Preliminary Design Chart for Slab Thickness Based on Two-Way Shear at an Edge

1-19
1.1
0.9
0.7
0.5
0.3
0.1
200 220 240 260 280 300 320 340 360 380 400
qu, psf
d/c1
c1
c1
˜2/2 + C1/2
˜1/2 + C1/2
A= (˜1/2 + C1/2)(˜2/2 + C2/2)
Panel
centerline
275
250
225
200
175
150
125
100
75
50
25
A
c2
1
=
Figure 1-10 Preliminary Design Chart for Slab Thickness Based on Two-Way Shear at a Corner

1-20
1.8.4 Footings
The required footing sizes can usually be obtained in a straightforward manner. In general, the base area is
determined by dividing the total service (unfactored) loads from the column by the allowable (safe) soil
pressure. In buildings without shearwalls, the maximum pressure due to the combination of gravity and wind
loads must also be checked. The required thickness may be obtained for either a reinforced or a plain footing
by using the appropriate simplified equation given in Chapter 7.
1.8.5 Fire Resistance
To insure adequate resistance to fire, minimum thickness and cover requirements are specified in building
codes as a function of the required fire resistance rating. Two hours is a typical rating for most members;
however, the local building code should be consulted for the ratings which apply to a specific project or
special portions of projects.
Member sizes that are necessary for structural requirements will usually satisfy the minimum requirements
for fire resistance as well (see Tables 10-1 and 10-2). Also, the cover requirements specified in ACI 7.7 will
provide at least a three-hour fire resistance rating for restrained floor members and columns (see Tables 10-3,
10-4, and 10-6).
It is important to check the fire resistance of a member immediately after a preliminary size has been obtained
based on structural requirements. Checking the fire resistance during the preliminary design stage eliminates
the possibility of having to completely redesign the member (or members) later on.
In the examples that appear in the subsequent chapters, the applicable fire resistance requirements tabulated in
Chapter 10 are checked for all members immediately after the preliminary sizes are obtained. The required fire
resistance ratings for both example buildings are listed in Section 1.7.
References
1.1 Building Code Requirements for Structural Concrete ACI 318-11 and Commentary—American Concrete
Institute, Famington Hills, Michigan, 2011.
1.2 Building Movements and Joints, EB086, Portland Cement Association, Skokie, Illinois 1982, 64 pp.
1.3 Concrete Floor Systems–Guide to Estimating and Economizing, SP041, Portland Cement Association,
Skokie, Illinois, 2000, 41 pp.
1.4 Long-Span Concrete Floor Systems, SP339, Portland Cement Association, Skokie, Illinois, 2000, 97 pp.
1.5 American Society of Civil Engineers Minimum Design Loads for Buildings and Other Structures,
ASCE/SEI 7-05, American Society of Civil Engineers, New York, N.Y., 2006.

1-21

2-1
Chapter 2
Simplified Frame Analysis
2.1 INTRODUCTION
The final design of the structural components in a building frame is based on maximum moment, shear, axial
load, torsion and/or other load effects, as generally determined by an elastic frame analysis (ACI 8.3).
For building frames of moderate size and height, preliminary and final designs will often be combined.
Preliminary sizing of members, prior to analysis, may be based on designer experience, design aids, or
simplified sizing expressions suggested in this book.
Analysis of a structural frame or other continuous construction is usually the most time consuming part of the
total design. For gravity load analysis of continuous one-way systems (beams and slabs), the approximate
moments and shears given by ACI 8.3.3 are satisfactory within the span and loading limitations stated.
For cases when ACI 8.3.3 is not applicable, a two-cycle moment distribution method is accurate enough.
The speed and accuracy of the method can greatly simplify the gravity load analysis of building frames with
usual types of construction, spans, and story heights. The method isolates one floor at a time and assumes that
the far ends of the upper and lower columns are fixed. This simplifying assumption is permitted by ACI 8.10.3.
For lateral load analysis of a sway frame, the Portal Method may be used. It offers a direct solution for the moments
and shears in the beams (or slabs) and columns, without having to know the member sizes or stiffnesses.
The simplified methods presented in this chapter for gravity load analysis and lateral load analysis
are considered to provide sufficiently accurate results for buildings of moderate size and height.
However, determinations of load effects by computer analysis or other design aids are equally applicable for
use with the simplified design procedures presented in subsequent chapters of this book.
2.2 LOADING
2.2.1 Service Loads
The first step in the frame analysis is the determination of design (service) loads and lateral forces (wind and
seismic) as called for in the general building code under which the project is to be designed and constructed.
For the purposes of this book, design live loads (and permissible reductions in live loads) and wind loads are
based on Minimum Design Loads for Buildings and Other Structures, ASCE/SEI 7-05.2.1 References to
specific ASCE Standard requirements are noted (ASCE 4.2 refers to ASCE/SEI 7-05, Section 4.2). For a
specific project, however, the governing general building code should be consulted for any variances from
ASCE/SEI 7-05.
Design dead loads include member self-weight, weight of fixed service equipment (plumbing, electrical, etc.) and,
where applicable, weight of built-in partitions. The latter may be accounted for by an equivalent uniform load of not
less than 20 psf, although this is not specifically defined in theASCE Standard (seeASCE Commentary Section 3.1.2).

Design live loads will depend on the intended use and occupancy of the portion or portions of the building
being designed. Live loads include loads due to movable objects and movable partitions temporarily supported
by the building during maintenance. In ASCE Table 4-1, uniformly distributed live loads range from 40 psf
for residential use to 250 psf for heavy manufacturing and warehouse storage. Portions of buildings, such
as library floors and file rooms, require substantially heavier live loads. Live loads on a roof include
maintenance equipment, workers, and materials. Also, snow loads, ponding of water, and special features, such
as landscaping, must be included where applicable.
Occasionally, concentrated live loads must be included; however, they are more likely to affect individual
supporting members and usually will not be included in the frame analysis (see ASCE 4.3).
2.2.2 Wind Loads
Wind pressure occurs as a result of transformation of the kinetic energy of the moving air when it is obstructed
by a building. The magnitude of pressure at a point depends on the velocity of the wind, the shape of the
building and the angle of incidence of the wind. Wind load effects on buildings are dynamic in nature. In case
of tall buildings, the dynamic nature of wind loads cannot be neglected. Tall buildings tend to deflect
appreciably and oscillate under wind effects. If a tall building is not properly designed for wind, the
oscillations may increase continuously leading to destruction. However, in many situations, such as buildings
of medium height and without irregularities, the dynamic nature of wind effects may be neglected. This allows
the designer to consider wind loads as statically applied pressure.
Design wind loads and procedures for design are usually given in the general building code having jurisdiction.
The calculation of wind loads in this publication is based on the procedure presented in ASCE/SEI 7-05 Standard,
Minimum Design Loads for Buildings and Other Structures2.1
. The ASCE/SEI 7-05 Standard provides three
methods to account for wind effects on buildings; simplified procedure (ASCE/SEI Sec. 6.4), analytical
procedure (ASCE/SEI Sec. 6.5) and wind tunnel procedure (ASCE/SEI Sec. 6.6). For the scope of buildings
covered in this publication, the analytical procedure is used.
In the ASCE/SE 7-05 analytical procedure the wind load is treated as static pressure. The calculations of the
pressure depend on the building size, geometry, importance, location, and openness. The magnitude of the
pressure varies with the height above ground. The procedure requires the determination of basic wind speed,
wind directionality factor, velocity pressure coefficient, gust effect factor and velocity pressure coefficient.
Table 2-1 summarizes ASCE/SEI 7-05 wind pressure parameters and the relevant Section in the Standard.
2-2

2-3
Chapter 2 • Simplified Frame Analysis
Symbol units ASCE 7 Equation / Notes
1 Basic wind speed V mph Figure 6-1
2 Occupancy category
I, II, III or IV
Table 1-1
3 Exposure category
B, C or D
6.5.6.3
4 Wind directionality
factor
Kd Table 6-4
5 Importance factor I Table 6-1 Based on Occupancy category from 2
6 Topographic factor Kzt 6.5.7 &
Figure 6-4
Kzt = (1+K1K2K3)2
where K1K2K3 are
determined from Figure 6-4
7 Gust effect factor G 6.5.8 For rigid structures may be assumed =
0.85
8 External pressure
coefficient
Cp Figure 6-6 For windward to be used with qz &
leeward to be used with qh depends on
the building ratio L/B
9 Velocity pressure
exposure coefficient
Kz for
different
heights
& Kh for
roof
6.5.6.6 &
Table 6-3
Depends on the exposure category
from 3 & the height
10 Velocity pressure qz psf 6.5.10 (Eq
6-15)
qz = 0.00256KzKztKdV2
I
11 Wind pressure pz psf 6.5.12.2 pz = qzGCp
Table 2-1 Wind Pressure Parameters

2.2.2.1 Example: Calculation of Wind Loads – Building #2
For illustration of the ASCE procedure, wind load calculations for the main wind-force resisting system of
building #2 (5-story flat plate) are summarized below.
Wind-force resisting system:
Alternate (1) - Slab and column framing with spandrel beams
Alternate (2) - Structural walls
(1) Wind load data
Assuming the building is classified as closed (ASCE 6.5.9) and located in Midwest in flat open terrain.
Basic wind speed V = 90 mph ASCE Figure 6-1
Occupancy category II ASCE Table 1-1
Exposure category C ASCE 6.5.6.3
Wind directionality factor Kd
= 0.85 ASCE Table 6-4
Importance factor I = 1 ASCE Table 6-1
Topographic factor Kzt
= 1 ASCE 6.5.7
Gust effect factor G = 0.85 ASCE 6.5.8
External pressure coefficient Cp
ASCE Figure 6-6
Windward - both directions Cp
= 0.8
Leeward - E-W direction (L/B = 120/60 = 2) Cp
= -0.3
N-S direction (L/B = 60/120 = 0.5) Cp
= -0.5
Velocity pressure exposure coefficients Kz
(ASCE 6.5.6.6) at various story heights are summarized in Table 2-1
Velocity pressure qz
= 0.00256Kz
Kzt
Kd
V2I ASCE 6.5.10
Design wind pressure pz = qzGCp ASCE 6.5.12.2
(2) Design wind pressure in the N-S Direction
Table 2-2 contains a summary of the design pressures calculations for wind in N-S direction.
Table 2-2 Design Pressures for N-S Direction
2-4
Level
Height
above
ground
level, z
(ft)
Kz qz
(psf)
Windward
design
pressure
qzGCp
(psf)
Leeward
design
pressure
qhGCp
(psf)
Total
design
pressure
(psf)
Roof 63 1.148 20.2 13.8 -8.6 22.4
4 51 1.098 19.4 13.2 -8.6 21.8
3 39 1.038 18.3 12.4 -8.6 21.0
2 27 0.961 16.9 11.5 -8.6 20.1
1 15 0.849 15.0 10.2 -8.6 18.8

(3) Wind loads in the N-S direction
The equivalent wind loads at each floor level are calculated as follows:
Alternate (1) Slab and column framing
Interior frame (24 ft bay width)
Roof = 22.4 ϫ 6.0 ϫ 24/1000 = 3.2 kips
4th = 21.8 ϫ 12 ϫ 24/1000 = 6.3 kips
3rd = 21.0 ϫ 12 ϫ 24/1000 = 6.1 kips
2nd = 20.1 ϫ 12 ϫ 24/1000 = 5.8 kips
1st = 18.8 ϫ 13.5 ϫ 24/1000 = 6.1 kips
Alternate (2) Structural walls
Total for entire building (121 ft width)
Roof = 22.4 ϫ 6.0 ϫ 121/1000 = 16.2 kips
4th = 21.8 ϫ 12 ϫ 121/1000 = 31.6 kips
3rd = 21.0 ϫ 12 ϫ 121/1000 = 30.6 kips
2nd = 20.1 ϫ 12 ϫ 121/1000 = 29.2 kips
1st = 18.8 ϫ 13.5 ϫ 121/1000 = 30.7 kips
(4) Wind loads in the E -W direction
Using the same procedure as for the N-S direction, the following wind loads are obtained for the E-W direction:
Alternate (1) Slab and column framing
Roof = 2.3 kips
4th = 4.4 kips
3rd = 4.2 kips
2nd = 4.0 kips
1st = 4.1 kips
Alternate (2) Structural walls
Total for entire building (61 ft width)
Roof = 6.9 kips
4th = 13.4 kips
3rd = 12.9 kips
2nd = 12.2 kips
1st = 12.6 kips
2-5

2-6
2.2.2.2 Example: Calculation of Wind Loads – Building #1
Wind load calculations for the main wind-force resisting system of Building #1 (3-story pan joist framing) are
summarized below.
Wind-force resisting system: Beam and column framing:
(1) Wind load data
Assuming the building is classified as closed (ASCE 6.5.9) and located in hurricane prone region.
Basic wind speed V = 145 mph ASCE Figure 6-1
Occupancy category II ASCE Table 1-1
Exposure category D ASCE 6.5.6.3
Wind directionality factor Kd
= 0.85 ASCE Table 6-4
Importance factor I = 1 ASCE Table 6-1
Topographic factor Kzt
= 1 ASCE 6.5.7
Gust effect factor G = 0.85 ASCE 6.5.8
External pressure coefficient Cp
ASCE Figure 6-6
Windward - both directions Cp
= 0.8
Leeward - E-W direction (L/B = 150/90 = 1.67) Cp
= -0.37
N-S direction (L/B = 90/150 = 0.6) Cp
= -0.5
Design pressures calculations for wind in N-S direction are summarized in Table 2-3
Table 2-3 Design Pressures for N-S Direction
(2) Summary of wind loads
N-S & E-W directions (conservatively use N-S wind loads in both directions):
Roof = 12 kips
2nd = 23.1 kips
1st = 21.7 kips
Note: The above loads were determined using design wind pressures computed at each floor level.
Level
Height
above
ground
level, z
(ft)
Kz qz
(psf)
Windward
design
pressure
qzGCp
(psf)
Leeward
design
pressure
qhGCp
(psf)
Total
design
pressure
(psf)
Roof 39 1.216 55.7 37.8 -23.7 61.5
2 27 1.141 51.9 35.3 -23.7 58.9
1 15 1.030 47.1 32.1 -23.7 55.7

2-7
2.2.3 Live Load Reduction for Columns, Beams, and Slabs
Most general building codes permit a reduction in live load for design of columns, beams and slabs to account
for the probability that the total floor area “influencing” the load on a member may not be fully loaded
simultaneously. Traditionally, the reduced amount of live load for which a member must be designed has been
based on a tributary floor area supported by that member. According to ASCE/SEI 7-05, the magnitude of live
load reduction is based on an influence area rather than a tributary area. The influence area is a function of the
tributary area for the structural member. The influence area for different structural members is calculated by
multiplying the tributary area for the member AT, by the coefficients KLL given in Table 2-4, see ASCE 4.8.
The reduced live load L per square foot of floor area supported by columns, beams, and two-way slabs having
an influence area (KLLAT) of more than 400 sq ft is:
ASCE (Eq. 4-1)
where Lo
is the unreduced design live load per square foot. The reduced live load cannot be taken less
than 50% for members supporting one floor, or less than 40% of the unit live load Lo
otherwise. For other
limitations on live load reduction, see ASCE 4.8.
Using the above expression for reduced live load, values of the reduction multiplier as a function of influence
area are given in Table 2-5.
Table 2-4 Live Load Element Factor KLL
The above limitations on permissible reduction of live loads are based on ASCE 4.8. The governing general
building code should be consulted for any difference in amount of reduction and type of members that may be
designed for a reduced live load.
L = L0
0.25 +
15
KLL
AT
⎛
⎝
⎜
⎞
⎠
⎟
Element KLL
Interior columns 4
Exterior column without cantilever slabs 4
Edge column with cantilever slabs 3
Corner columns with cantilever slabs 2
Edge beams without cantilever slabs 2
Interior beams 2
All other members not identified above including:
Edge beams with cantilever slabs
Cantilever beams
Two-way slabs
1

Table 2-5 Reduction Multiplier (RM) for Live Load =
2.2.3.1 Example: Live Load Reductions for Building #2
For illustration, typical influence areas for the columns and the end shear walls of Building #2 (5-story flat
plate) are shown in Fig. 2-1. Corresponding live load reduction multipliers are listed in Table 2-6.
Figure 2-1 Typical Influence Areas
2-8
Influence Area
KLLAT RM
Influence Area
KLLAT RM
400a
1.000 5600 0.450
800 0.780 6000 0.444
1200 0.683 6400 0.438
1600 0.625 6800 0.432
2000 0.585 7200 0.427
2400 0.556 7600 0.422
2800 0.533 8000 0.418
3200 0.515 8400 0.414
3600 0.500b
8800 0.410
4000 0.487 9200 0.406
4800 0.467 10000 0.400c
5200 0.458
a
No live load reduction is permitted for influence area less than 400 sq ft.
b
Maximum reduction permitted for members supporting one floor only.
c
Maximum absolute reduction.
0.25 +
15
KLL
AT
⎛
⎝
⎜
⎞
⎠
⎟

For the interior columns, the reduced live load is L = 0.42Lo
at the first story (AT
= 4 bay areas ϫ 4 stories =
20 ϫ 24 ϫ 4 ϫ 4 = 7680 sq ft). The two-way slab may be designed with an RM = 0.94 (AT
= 480 sq ft for one
bay area). Shear strength around the interior columns is designed for an RM = 0.59 (AT
= 1920 sq ft for 4 bay
areas), and around an edge column for an RM = 0.73 (AT
= 960 sq ft for 2 bay areas). Spandrel beams could
be designed for an RM = 0.94 (one bay area). If the floor system were a two-way slab with beams between
columns, the interior beams would qualify for an RM = 0.73 (2 bay areas).
2.2.4 Factored Loads
The strength design method using factored loads to proportion members is used exclusively in this book.
The design (service) loads must be increased by specified load factors (ACI 9.2), and factored loads must be
combined in load combinations depending on the types of loads being considered. The method requires that the
design strength of a member at any section should equal or exceed the required strength calculated by the code-
specified factored load combinations. In general,
Design Strength ≥ Required Strength
or
Strength Reduction Factor (φ) ϫ Nominal Strength ≥ Load Factor ϫ Service Load Effects
All structural members and sections must be proportioned to meet the above criterion under the most critical
load combination for all possible load effects (flexure, axial load, shear, etc.).
For structures subjected to dead, live, wind, and earthquake loads only, the ACI applicable load combinations
are summarized in Table 2-7:
The strength reduction factors are listed below:
Tension-controlled sections* 0.90
Compression-controlled sections*
Members with spiral reinforcement conforming to 10.9.3 0.75
Other reinforced members 0.65
Shear and torsion 0.75
Bearing on concrete 0.65
2-9
Table 2-6 Reduction Multiplier (RM) for Live Loads, Building #2
Interior Columns Edge Columns Corner Columns End Shear Wall Units
Story KLLAT (ft2
) RM KLLAT (ft2
) RM KLLAT (ft2
) RM KLLAT (ft2
) RM
5 (roof) * * * *
4 1920 0.59 960 0.73 480 0.94 1440 0.65
3 3840 0.49 1920 0.59 960 0.73 2880 0.53
2 5760 0.45 2880 0.53 1440 0.65 4320 0.48
1 7680 0.42 3840 0.49 1920 0.59 5760 0.48
*No reduction permitted for roof live load (ASCE 4.8.2); the roof should not be included in the influence areas of the
floors below.
* For definition of tension controlled, compression controlled sections and strength reduction for transition sections, see Chapters 3 and 5.

2-10
Table 2-7(a) ACI Load Combinations for Building Subjected for Dead, Live Wind, and Earthquake Loads
(W is based on service-level wind loads)
Table 2-7(b) ACI Load Combinations for Building Subjected for Dead, Live Wind, and Earthquake Loads
(W is based on strength-level wind loads)
The ACI Code permits the load factor applied to the live load L to be reduced to 0.5 in Eq. (9-3) to (9-5)
(ACI 9.2.1). This reduction cannot be applied for garages, areas occupied as places of public assembly, and all
areas where L is greater than 100 lb/ft2
. It is important to point out that this reduction in the load factor can be
applied in addition to the live load reduction described in Section 2.2.3.
For design of beams and slabs, the factored load combinations used most often are:
U = 1.4D ACI Eq. (9-1)
U = 1.2D + 1.6L ACI Eq. (9-2)
ACI Eq. (9-1) seldom controls—only when the live load to dead load ratio (L/D) is less than 0.125.
For a frame analysis with live load applied only to a portion of the structure, i.e., alternate spans (ACI 8.11),
the factored loads to be applied would be computed separately using the appropriate load factor for each load.
ACI
Equations
Garages, places of public assembly
and all areas where L is greater than
100 lb/ft
2
Other buildings
Eq. (9-1) U = 1.4D U = 1.4D
Eq. (9-2) U = 1.2D + 1.6L + 0.5Lr U = 1.2D + 1.6L + 0.5Lr
Eq. (9-3) U = 1.2D + 1.6Lr + 1.0L
U = 1.2D + 1.6Lr + 0.8W
U = 1.2D + 1.6Lr + 0.5L
U = 1.2D + 1.6Lr + 0.8W
Eq. (9-4) U = 1.2D + 1.6W + 1.0L + 0.5Lr U = 1.2D + 1.6W + 0.5L + 0.5Lr
Eq. (9-5) U = 1.2D + 1.0E + 1.0L U = 1.2D + 1.0E + 0.5L
Eq. (9-6) U = 0.9D + 1.6W U = 0.9D + 1.6W
Eq. (9-7) U = 0.9D + 1.0E U = 0.9D + 1.0E
ACI
Equations
Garages, places of public assembly
and all areas where L is greater than
100 lb/ft
2
Other buildings
Eq. (9-1) U = 1.4D U = 1.4D
Eq. (9-2) U = 1.2D + 1.6L + 0.5Lr U = 1.2D + 1.6L + 0.5Lr
Eq. (9-3) U = 1.2D + 1.6Lr + 1.0L
U = 1.2D + 1.6Lr + 0.5W
U = 1.2D + 1.6Lr + 0.5L
U = 1.2D + 1.6Lr + 0.5W
Eq. (9-4) U = 1.2D + 1.0W + 1.0L + 0.5Lr U = 1.2D + 1.0W + 0.5L + 0.5Lr
Eq. (9-5) U = 1.2D + 1.0E + 1.0L U = 1.2D + 1.0E + 0.5L
Eq. (9-6) U = 0.9D + 1.0W U = 0.9D + 1.0W
Eq. (9-7) U = 0.9D + 1.0E U = 0.9D + 1.0E
D = dead loads, or related internal moments and forces
L = live loads, or related internal moments and forces
Lr = roof live load, or related internal moments and forces
U = required strength to resist factored loads or related internal moments and forces
W = wind load, or related internal moments and forces
E = earthquake loads, or related internal moments and forces

2-11
The designer has the choice of multiplying the service loads by the load factors before computing the factored
load effects (moments, shears, etc.), or computing the effects from the service loads and then multiplying the
effects by the load factors. For example, in the computation of bending moment for dead and live loads
U = 1.2D + 1.6L, the designer may (1) determine wu
= 1.2 wd
+ 1.6 w˜
and then compute the factored moments
using wu
; or (2) compute the dead and live load moments using service loads and then determine the factored
moments as Mu
= 1.2 Md
+ 1.6 M˜
. Both analysis procedures yield the same answer. It is important to note that
the second alternative is much more general than the first; thus, it is more suitable for computer analysis, espe-
cially when more than one load combination must be investigated.
2.3 FRAME ANALYSIS BY COEFFICIENTS
The ACI Code provides a simplified method of analysis for both one-way construction (ACI 8.3.3) and
two-way construction (ACI 13.6). Both simplified methods yield moments and shears based on coefficients.
Each method will give satisfactory results within the span and loading limitations stated in Chapter 1.
The direct design method for two-way slabs is discussed in Chapter 4.
2.3.1 Continuous Beams and One-Way Slabs
When beams and one-way slabs are part of a frame or continuous construction, ACI 8.3.3 provides approximate
moment and shear coefficients for gravity load analysis. The approximate coefficients may be used as long as
all of the conditions illustrated in Fig. 2-2 are satisfied: (1) There must be two or more spans, approximately
equal in length, with the longer of two adjacent spans not exceeding the shorter by more than 20 percent;
(2) loads must be uniformly distributed, with the service live load not more than 3 times the dead load
(L/D ″ 3); and (3) members must have uniform cross section throughout the span. Also, no redistribution of
moments is permitted (ACI 8.4). The moment coefficients defined in ACI 8.3.3 are shown in Figs. 2-3 through
2-6. In all cases, the shear in end span members at the interior support is taken equal to 1.15wu
ñ
/2. The shear
at all other supports is wu
/2 (see Fig. 2-7). wu
ñ
is the combined factored load for dead and live loads,
wu
= 1.2wd
+ 1.6 w˜
. For beams, wu
is the uniformly distributed load per unit length of beam (plf), and the
coefficients yield total moments and shears on the beam. For one-way slabs, wu
is the uniformly distributed load
per unit area of slab (psf), and the moments and shears are for slab strips one foot in width. The span length ñ
is defined as the clear span of the beam or slab. For negative moment at a support with unequal adjacent spans,
ñ is the average of the adjacent clear spans. Support moments and shears are at the faces of supports.
Figure 2-2 Conditions for Analysis by Coefficients (ACI 8.3.3)
Prismatic
Members
n≤ 1.2 n
Uniformly Distributed Load L / D ≤ 3( )

2-12
Simple
Support
ñ
Wu˜2
n
11
Wu˜2
n
16
Wu˜2
n
14
ñ ñ
End Span
Integral
with
Support
Interior SpanEnd Span
Simple
Support
ñ
Wu˜2
n
12
Wu˜2
n
12
ñ ñ
Integral
with
Support
Wu˜2
n
12
Wu˜2
n
12
Wu˜2
n
12
Simple
Support
ñ
Wuñ
2
ñ ñ
Integral
with
Support
Wuñ
2
Wuñ
2
Wuñ
2
End SpanInterior SpanEnd Span
1.15Wuñ
2
1.15Wuñ
2
ñ ñ ñ
Wu˜2
n
12
Wu˜2
n
12
Wu˜2
n
12
Wu˜2
n
12
Wu˜2
n
12
Wu˜2
n
12
Simple
Support
ñ
Wu˜2
n*
10
Wu˜2
n
11
ñ ñ
Integral
with
Support
Wu˜2
n
11 10
Wu˜2
n
24
Wu˜2
n* Spandrel
Support
Column
Support
Wu˜2
n
16
*Wu˜2
n
9
(2 spans)
Figure 2-7 End Shears—All Cases
Figure 2-3 Positive Moments—All Cases
Figure 2-4 Negative Moments—Beams and Slabs
Figure 2-5 Negative Moments—Slabs with spans < 10 ft
Figure 2-6 Negative Moments—Beams with Stiff Columns

2-13
2.3.2 Example: Frame Analysis by Coefficients
Determine factored moments and shears for the joists of the standard pan joist floor system of Building #1
(Alternate (1)) using the approximate moment and shear coefficients ofACI 8.3.3. Joists are spaced at 3 ft on center.
(1) Data: Width of spandrel beam = 20 in.
Width of interior beams = 36 in.
Floors: LL = 60 psf
DL = 130 psf
wu = 1.2 (130) + 1.6 (60) = 252 psf ϫ 3 ft = 756 plf
(2) Factored moments and shears using the coefficients from Figs. 2-3, 2-4, and 2-7 are summarized in Fig. 2-8.
15'-0"30'-0"30'-0"
8" 1'-0" ñ = 27.5' 3'-0" ñ = 27.0' 3'-0" ñ = 27.0'
Sym. aboutCL
*Average of adjacent clear spans
Total Load
Coefficient
from
Fig. 2-7
Shear Vu
wuñ
2
Coefficient
from
Fig. 2-3
Pos. Mu
ñ
wuñ
2
Coefficient
from
Fig. 2-3
Neg. Mu
wuñ = 0.76 x 27.5
= 20.9 kips
0.76 x 27.0 = 20.5 kips 0.76 x 27.0 = 20.5 kips
10.5 kips 12 kips 10.3 kips 10.3 kips 10.3 kips
1/2
1/2
1/2
1/2
574.8 ft-kips 554 ft-kips 554 ft-kips
1/14
1/16
1/16
1.15/2
41.1 ft-kips 34.6 ft-kips 34.6 ft-kips
27.5 ft 27.25 ft* 27.25 ft* 27.0 ft 27.0 ft
574.8 ft-kips 564.3 ft-kips 554 ft-kips564.3 ft-kips 554 ft-kips
1/24
1/11
1/11
1/11
1/10
24 ft-kips 51.3 ft-kips 50.4 ft-kips56.3 ft-kips 50.4 ft-kips
Figure 2-8 Factored Moments and Shears for the Joist Floor System of Building #1 (Alternate 1)

2-14
2.4 FRAME ANALYSIS BY ANALYTICAL METHODS
For continuous beams and one-way slabs not meeting the limitations of ACI 8.3.3 for analysis by coefficients,
an elastic frame analysis must be used. Approximate methods of frame analysis are permitted by ACI 8.3.2 for
“usual” types of buildings. Simplifying assumptions on member stiffnesses, span lengths, and arrangement of
live load are given in ACI 8.7 through 8.11.
2.4.1 Stiffness
The relative stiffnesses of frame members must be established regardless of the analytical method used. Any
reasonable consistent procedure for determining stiffnesses of columns, walls, beams, and slabs is permitted by
ACI 8.7.
The selection of stiffness factors will be considerably simplified by the use of Tables 2-8 and 2-9. The stiffness
factors are based on gross section properties (neglecting any reinforcement) and should yield satisfactory results for
buildings within the size and height range addressed in this book. In most cases where an analytical procedure is
required, stiffness of T-beam sections will be needed. The relative stiffness values K given in Table 2-8 allow for
the effect of the flange by doubling the moment of inertia of the web section (bw
h). For values of hf
/h between
0.2 and 0.4, the multiplier of 2 corresponds closely to a flange width equal to six times the web width. This is
considered a reasonable allowance for most T-beams.2.2
For rectangular beam sections; the tabulated values should
be divided by 2. Table 2-9 gives relative stiffness values K for column sections. It should be noted that column
stiffness is quite sensitive to changes in column size. The initial judicious selection of column size and uniformity
from floor to floor is, therefore, critical in minimizing the need for successive analyses.
As is customary for ordinary building frames, torsional stiffness of transverse beams is not considered in the
analysis. For those unusual cases where equilibrium torsion is involved, a more exact procedure may be necessary.

2-15
2.4.3 Design Moments
When determining moments in frames or continuous construction, the span length shall be taken as the
distance center-to-center of supports (ACI 8.9.2). Moments at faces of supports may be used for member design
purposes (ACI 8.9.3). Reference 2.2 provides a simple procedure for reducing the centerline moments to face
of support moments, which includes a correction for the increased end moments in the beam due to the
restraining effect of the column between face and centerline of support. Figure 2-10 illustrates this correction.
For beams and slabs subjected to uniform loads, negative moments from the frame analysis can be reduced by
wu˜2
a/6. A companion reduction in the positive moment of wu˜2
a/12 can also be made.
A B C D
A B C D
A B C D
(2) Loading pattern for negative moment at support B
(3) Loading pattern for positive moment in span BC
wd + w˜
wdwd
wd + w˜
(1) Loading pattern for negative moment at support A
and positive moment in span AB
wd + w˜
wd
wd
Figure 2-9 Partial Frame Analysis for Gravity Loading
2.4.2 Arrangement of Live Load
According to ACI 8.11.1, it is permissible to assume that for gravity load analysis, the live load is applied only to
the floor or roof under consideration, with the far ends of the columns assumed fixed. In the usual case where the
exact loading pattern is not known, the most demanding sets of design forces must be investigated. Figure 2-9
illustrates the loading patterns that should be considered for a three-span frame.

2-16
Table 2-8 Beam Stiffness Factors
h
hf
bw
Moment of inertia, excluding
overhanging flanges:
Moment of inertia of T-section =~ 2I
=
bwh
3
12
=
2
10
I
I
˜
K*
Values of K for T-beams
Span of beam, l (ft)** Span of beam, l (ft)**
h bw I 8 10 12 14 16 20 24 30 h bw I 8 10 12 14 16 20 24 30
6 256 6 5 4 4 3 3 2 2 8 9216 230 185 155 130 115 90 75 60
8 341 9 7 6 5 4 3 3 2 10 11520 290 230 190 165 145 115 95 75
10 427 11 9 7 6 5 4 4 3 11.5 13248 330 265 220 190 165 130 110 90
8 11.5 491 12 10 8 7 6 5 4 3 24 13 14976 375 300 250 215 185 150 125 100
13 555 14 11 9 8 7 6 5 4 15 17280 430 345 290 245 215 175 145 115
15 640 16 13 11 9 8 6 5 4 17 19584 490 390 325 280 245 195 165 130
17 725 18 15 12 10 9 7 6 5 19 21888 545 440 365 315 275 220 180 145
19 811 20 16 14 12 10 8 7 5 21 24192 605 485 405 345 300 240 200 160
6 500 13 10 8 7 6 5 4 3 8 11717 295 235 195 165 145 115 100 80
8 667 17 13 11 10 8 7 6 4 10 14647 365 295 245 210 185 145 120 100
10 833 21 17 14 12 10 8 7 6 11.5 16844 420 335 280 240 210 170 140 110
10 11.5 958 24 19 16 14 12 10 8 6 26 13 19041 475 380 315 270 240 190 160 125
13 1083 27 22 18 15 14 11 9 7 15 21970 550 440 365 315 275 220 185 145
15 1250 31 25 21 18 16 13 10 8 17 24899 620 500 415 355 310 250 205 165
17 1417 35 28 24 20 18 14 12 9 19 27892 695 555 465 400 350 280 230 185
19 1583 40 32 26 23 20 16 13 11 21 30758 770 615 515 440 385 310 255 205
6 864 22 17 14 12 11 9 7 6 8 14635 365 295 245 210 185 145 120 100
8 1152 29 23 19 16 14 12 10 8 10 18293 455 365 305 260 230 185 150 120
10 1440 36 29 24 21 18 14 12 10 11.5 21037 525 420 350 300 265 210 175 140
12 11.5 1656 41 33 28 24 21 17 14 11 28 13 23781 595 475 395 340 295 240 200 160
13 1872 47 37 31 27 23 19 16 12 15 27440 685 550 455 390 345 275 230 185
15 2160 54 43 36 31 27 22 18 14 17 31099 775 620 520 445 390 310 260 205
17 2448 61 49 41 35 31 25 20 16 19 34757 870 695 580 495 435 350 290 230
19 2736 68 55 46 39 34 27 23 18 21 38416 960 770 640 550 480 385 320 255
6 1372 34 27 23 20 17 14 11 9 8 18000 450 360 300 255 225 180 150 120
8 1829 46 37 30 26 23 18 15 12 10 22500 565 450 375 320 280 225 190 150
10 2287 57 46 38 33 29 23 19 15 11.5 25875 645 520 430 370 325 260 215 175
14 11.5 2630 66 53 44 38 33 26 22 18 30 13 29250 730 585 490 420 365 295 245 195
13 2973 74 59 50 42 37 30 25 20 15 33750 845 675 565 480 420 340 280 225
15 3430 86 69 57 49 43 34 29 23 17 38250 955 765 640 545 480 385 320 255
17 3887 97 78 65 56 49 39 32 26 19 42750 1070 855 715 610 535 430 355 285
19 4345 109 87 72 62 54 43 36 29 21 47250 1180 945 790 675 590 475 395 315
6 2048 51 41 34 29 26 20 17 14 8 31104 780 620 520 445 390 310 260 205
8 2731 68 55 46 39 34 27 23 18 10 38880 970 780 650 555 485 390 325 260
10 3413 85 68 57 49 43 34 28 23 11.5 44712 1120 895 745 640 560 445 375 300
16 11.5 3925 98 79 65 56 49 39 33 26 36 13 50544 1260 1010 840 720 630 505 420 335
13 4437 111 89 74 63 55 44 37 30 15 58320 1460 1170 970 835 730 585 485 390
15 5120 128 102 85 73 64 51 43 34 17 66096 1650 1320 1100 945 825 660 550 440
17 5803 145 116 97 83 73 58 48 39 19 73872 1850 1480 1230 1060 925 740 615 490
19 6485 162 130 108 93 81 65 54 43 21 81648 2040 1630 1360 1170 1020 815 680 545
6 2916 73 58 49 42 36 29 24 19 8 49392 1230 990 825 705 615 495 410 330
8 3888 97 78 65 56 49 39 32 26 10 61740 1540 1230 1030 880 770 615 515 410
10 4860 122 97 81 69 61 49 41 32 11.5 71001 1780 1420 1180 1010 890 710 590 475
18 11.5 5589 140 112 93 80 70 56 47 37 42 13 80262 2010 1610 1340 1150 1000 805 670 535
13 6318 158 126 105 90 79 63 53 42 15 92610 2320 1850 1540 1320 1160 925 770 615
15 7290 182 146 122 104 91 73 61 49 17 104958 2620 2100 1750 1500 1310 1050 875 700
17 8262 207 165 138 118 103 83 69 55 19 117306 2930 2350 1950 1680 1470 1170 975 780
19 9234 231 185 154 132 115 92 77 62 21 129654 3240 2590 2160 1850 1620 1300 1080 865
6 4000 100 80 67 57 50 40 33 27 8 73728 1840 1470 1230 1050 920 735 615 490
8 5333 133 107 89 76 67 53 44 36 10 92160 2300 1840 1540 1320 1150 920 770 615
10 6667 167 133 111 95 83 67 56 44 11.5 105984 2650 2120 1770 1510 1320 1060 885 705
20 11.5 7667 192 153 128 110 96 77 64 51 48 13 119808 3000 2400 2000 1710 1500 1200 1000 800
13 8667 217 173 144 124 108 87 72 58 15 138240 3460 2760 2300 1970 1730 1380 1150 920
15 10000 250 200 167 143 125 100 83 67 17 156672 3920 3130 2610 2240 1960 1570 1310 1040
17 11333 283 227 189 162 142 113 94 76 19 175104 4380 3500 2920 2500 2190 1750 1460 1170
19 12667 317 253 211 181 158 127 106 84 21 193536 4840 3870 3230 2760 2420 1940 1610 1290
6 5324 133 106 89 76 67 53 44 36 8 104976 2620 2100 1750 1500 1310 1050 875 700
8 7099 177 142 118 101 89 71 59 47 10 131220 3280 2620 2190 1880 1640 1310 1090 875
10 8873 222 177 148 127 111 89 74 59 11.5 150903 3770 3020 2510 2160 1890 1510 1260 1010
22 11.5 10204 255 204 170 146 128 102 85 68 54 13 170586 4260 3410 2840 2440 2130 1710 1420 1140
13 11535 288 231 192 165 144 115 96 77 15 196830 4920 3490 3280 2810 2460 1970 1640 1310
15 13310 333 266 222 190 166 133 111 89 17 223074 5580 4460 3720 3190 2790 2230 1860 1490
17 15085 377 302 251 215 189 151 126 101 19 249318 6230 4990 4160 3560 3120 2490 2080 1660
19 16859 421 337 281 241 211 169 141 112 21 275562 6890 5510 4590 3940 3440 2760 2300 1840
*Coefficient 10 introduced to reduce magnitude of relative stiffness values
**Center-to-center distance between supports

2-17
Table 2-9 Beam Stiffness Factors
h
b
Values of K for T-beams
Span of beam, (ft)** Span of beam, (ft)**
h b I 8 9 10 11 12 14 16 20 h b I 8 9 10 11 12 14 16 20
10 427 5 5 4 4 4 3 3 2 12 13824 175 155 140 125 115 100 85 70
12 512 6 6 5 5 4 4 3 3 14 16128 200 180 160 145 135 115 100 80
14 597 7 7 6 5 5 4 4 3 18 20738 260 230 205 190 175 150 130 105
8 18 766 10 9 8 7 6 5 5 4 24 22 25344 315 280 255 230 210 180 160 125
22 939 12 10 9 9 8 7 6 5 26 29952 375 335 300 270 250 215 185 150
26 1109 14 12 11 10 9 8 7 6 30 34560 430 385 345 315 290 245 215 175
30 1280 16 14 13 12 11 9 8 6 36 41472 520 460 415 375 345 295 260 205
36 1536 19 17 15 14 13 11 10 8 42 48384 605 540 485 440 405 345 300 240
10 833 10 9 8 8 7 6 5 4 12 17576 220 195 175 160 145 125 110 90
12 1000 13 11 10 9 8 7 6 5 14 20505 255 230 205 185 170 145 130 105
14 1167 15 13 12 11 10 8 7 6 18 26364 330 295 265 240 220 190 165 130
10 18 1500 19 17 15 14 13 11 9 8 26 22 32223 405 360 320 295 270 230 200 160
22 1833 23 20 18 17 15 13 11 9 26 38081 475 425 380 345 315 270 240 190
26 2167 27 24 22 20 18 16 14 11 30 43940 550 490 440 400 365 315 275 220
30 2500 31 28 25 23 21 18 16 13 36 52728 660 585 525 480 440 375 330 265
36 3000 38 33 30 27 25 21 19 15 42 61516 770 685 615 560 515 440 385 310
10 1440 18 16 14 13 12 10 9 7 12 21952 275 245 220 200 185 155 135 110
12 1728 22 19 17 16 14 12 11 9 14 25611 320 285 255 235 215 185 160 130
14 2016 25 22 20 18 17 14 13 10 18 32928 410 365 330 300 275 235 205 165
12 18 2592 32 29 26 24 22 19 16 13 28 22 40245 505 445 400 365 335 285 250 200
22 3168 40 35 32 29 26 23 20 16 26 47563 595 530 475 430 395 340 295 240
26 3744 47 42 37 34 31 27 23 19 30 54880 685 610 550 500 455 390 345 275
30 4320 54 48 43 39 36 31 27 22 36 65856 825 730 660 600 550 470 410 330
36 5184 65 58 52 47 43 37 32 26 42 76832 960 855 770 700 640 550 480 385
10 2287 29 25 23 21 19 16 14 11 12 27000 340 300 270 245 225 195 170 135
12 2744 34 30 27 25 23 20 17 14 14 31500 395 350 315 285 265 225 195 160
14 3201 40 36 32 29 27 23 20 16 18 40500 505 450 405 370 340 290 255 205
14 18 4116 51 46 41 37 34 29 26 21 30 22 49500 620 550 495 450 415 355 310 250
22 5031 63 56 50 46 42 36 31 25 26 58500 730 650 585 530 490 420 365 295
26 5945 74 66 59 54 50 42 37 30 30 67500 845 750 675 615 565 480 420 340
30 6860 86 76 69 62 57 49 43 34 36 81000 1010 900 810 735 675 580 505 405
36 8232 103 91 82 75 69 59 51 41 42 94500 1180 1050 945 860 790 675 590 475
10 3413 43 38 34 31 28 24 21 17 12 32768 410 365 330 300 275 235 205 165
12 4096 51 46 41 37 34 29 26 20 14 38229 480 425 380 350 320 275 240 190
14 4779 60 53 48 43 40 34 30 24 18 49152 615 545 490 445 410 350 305 245
16 18 6144 77 68 61 56 51 44 38 31 32 22 60075 750 670 600 545 500 430 375 600
22 7509 94 83 75 68 63 54 47 38 26 70997 885 790 710 645 590 505 445 355
26 8875 111 99 89 81 74 63 55 44 30 81920 1020 910 820 745 685 585 510 410
30 10240 128 114 102 93 85 73 64 51 36 98304 1230 1090 985 895 820 700 615 490
36 12288 154 137 123 112 102 88 77 61 42 114688 1430 1270 1150 1040 955 820 715 575
10 4860 61 54 49 44 41 35 30 24 12 39304 490 435 395 355 330 280 245 195
12 5832 73 65 58 53 49 42 36 29 14 45855 575 510 460 415 380 330 285 230
14 6804 85 76 68 62 57 49 43 34 18 58956 735 655 590 535 490 420 370 295
18 18 8748 109 97 87 80 73 62 55 44 34 22 72057 900 800 720 655 600 515 450 360
22 10692 134 119 107 97 89 76 67 53 26 85159 1060 945 850 775 710 610 530 425
26 12636 158 140 126 115 105 90 79 63 30 98260 1230 1090 985 895 820 700 615 490
30 14580 182 162 146 133 122 104 91 73 36 117912 1470 1310 1180 1070 980 840 735 590
36 17496 219 194 175 159 146 125 109 87 42 137564 1720 1530 1380 1250 1150 985 860 690
10 6667 83 74 67 61 56 48 42 33 12 46656 585 520 465 425 390 335 290 235
12 8000 100 89 80 73 67 57 50 40 14 54432 680 605 545 495 455 390 340 270
14 9333 117 104 93 85 78 67 58 47 18 69984 875 780 700 635 585 500 435 350
20 18 12000 150 133 120 109 100 86 75 60 36 22 85536 1070 950 855 780 715 610 535 430
22 14667 183 163 147 133 122 105 92 73 26 101088 1260 1120 1010 920 840 720 630 505
26 17333 217 193 173 158 144 124 108 87 30 116640 1460 1300 1170 1060 970 835 730 585
30 20000 250 222 200 182 167 143 125 100 36 139968 1750 1560 1400 1270 1170 1000 875 700
36 24000 300 267 240 218 200 171 150 120 42 163296 2040 1810 1630 1480 1360 1170 1020 815
10 8873 111 99 89 81 74 63 55 44 12 54872 685 610 550 500 460 390 345 275
12 10648 133 118 106 97 89 76 67 53 14 64017 800 710 640 580 535 455 400 320
14 12422 155 138 124 113 104 89 78 62 18 82308 1030 915 825 750 685 590 515 410
22 18 15972 200 177 160 145 133 114 100 80 38 22 100599 1260 1120 1010 915 840 720 630 505
22 19521 244 217 195 177 163 139 122 98 26 118889 1490 1320 1190 1080 990 850 745 595
26 23071 288 256 231 210 192 165 144 115 30 137180 1710 1520 1370 1250 1140 980 855 685
30 26620 333 296 266 242 222 190 166 133 36 164616 2060 1830 1650 1500 1370 1180 1030 825
36 31944 399 355 319 290 266 228 200 160 42 192052 2400 2130 1920 1750 1600 1370 1200 960
*Coefficient 10 introduced to reduce magnitude of relative stiffness values
**Center-to-center distance between supports
I =
bh3
12
K*
=
I
10 c

2-18
Figure 2-10 Correction Factors for Span Moments 2.2
2.4.4 Two-Cycle Moment Distribution Analysis for Gravity Loading
Reference 2.2 presents a simplified two-cycle method of moment distribution for ordinary building frames.
The method meets the requirements for an elastic analysis called for in ACI 8.3 with the simplifying assump-
tions of ACI 8.6 through 8.9.
The speed and accuracy of the two-cycle method will be of great assistance to designers. For an in-depth
discussion of the principles involved, the reader is directed to Reference 2.2.
2.5 COLUMNS
In general, columns must be designed to resist the axial loads and maximum moments from the combination
of gravity and lateral loading.
For interior columns supporting two-way construction, the maximum column moments due to gravity loading can be
obtained by using ACI Eq. (13-7) (unless a general analysis is made to evaluate gravity load moments from alternate
span loading). With the same dead load on adjacent spans, this equation can be written in the following form:
Mu
= 0.07 qDu
2
n
− ʹ2
n( )+ 0.5qLu
2
n
⎡
⎣
⎤
⎦ 2
Modified
Computed
from analysis
Face of col.
column
span
A
BC
CL
uw a˜2
4
uw a˜2
6
uw a˜2
12
CL
uw a˜2
12
c
2
a˜
2
=
(A) = Theoretical moment
(B) = Computed moment ignoring stiffening effect of column support
(C) = Modified moment at face of column
= uniformly distributed factored load (plf)
= span length center-to-center of supports
= width of column support
= c/˜
˜
CL
CL
uw
c
a

2-19
where:
qDu
= uniformly distributed factored dead load, psf
qLu
= uniformly distributed factored live load (including any live load reduction; see
Section 2.2.2), psf
˜n
= clear span length of longer adjacent span, ft
= clear span length of shorter adjacent span, ft
˜2
= length of span transverse to ˜n
and , measured center-to-center of supports, ft
For equal adjacent spans, this equation further reduces to:
The factored moment Mu
can then be distributed to the columns above and below the floor in proportion to their
stiffnesses. Since the columns will usually have the same cross-sectional area above and below the floor under
consideration, the moment will be distributed according to the inverse of the column lengths.
2.6 LATERAL LOAD ANALYSIS
For frames without shear walls, the lateral load effects must be resisted by the “sway” frame. For low-to-
moderate height buildings, lateral load analysis of a sway frame can be performed by either of two simplified
methods: the Portal Method or the Joint Coefficient Method. Both methods can be considered to satisfy the
elastic frame analysis requirements of the code (ACI 8.3). The two methods differ in overall approach.
The Portal Method considers a vertical slice through the entire building along each row of column lines.
The method is well suited to the range of building size and height considered in this book, particularly to build-
ings with a regular rectangular floor plan. The Joint Coefficient Method considers a horizontal slice through
the entire building, one floor at a time. The method can accommodate irregular floor plans, and provision is
made to adjust for a lateral loading that is eccentric to the centroid of all joint coefficients (centroid of
resistance). The Joint Coefficient Method considers member stiffnesses, whereas the Portal Method does not.
The Portal Method is presented in this book because of its simplicity and its intended application to buildings
of regular shape. If a building of irregular floor plan is encountered, the designer is directed to Reference 2.2
for details of the Joint Coefficient Method.
2.6.1 Portal Method
The Portal Method considers a two-dimensional frame consisting of a line of columns and their connecting
horizontal members (slab-beams), with each frame extending the full height of the building. The frame is
considered to be a series of portal units. Each portal unit consists of two story-high columns with connecting
slab-beams. Points of contraflexure are assumed at mid-length of beams and mid-height of columns. Figure 2-11
illustrates the portal unit concept applied to the top story of a building frame, with each portal unit shown
separated (but acting together).
The lateral load W is divided equally between the three portal units. The shear in the interior columns is twice that
in the end columns. In general, the magnitude of shear in the end column is W/2n, and in an interior column it is
W/n, where n is the number of bays. For the case shown with equal spans, axial load occurs only in the end columns
since the combined tension and compression due to the portal effect results in zero axial loads in the interior
ʹn
Mu
= 0.07 0.5qLu
2
n( ) 2
= 0.035qLu
2
n 2
ʹn

2-20
columns. Under the assumptions of this method, however, a frame configuration with unequal spans will have axial
load in those columns between the unequal spans, as well as in the end columns. The general term for axial load in
the end columns in a frame of n bays with unequal spans is:
= length of bay n
The axial load in the first interior column is:
and, in the second interior column:
Column moments are determined by multiplying the column shear with one-half the column height. Thus, for
joint B in Fig. 2-11, the column moment is (W/3) (h/2) = Wh/6. The column moment Wh/6 must be balanced
by equal moments in beams BA and BC, as shown in Fig. 2-12.
Note that the balancing moment is divided equally between the horizontal members without considering their
relative stiffnesses. The shear in beam AB or BC is determined by dividing the beam end moment by one-half
the beam length, (Wh/12)(˜/2) = Wh/6˜ .
The process is continued throughout the frame taking into account the story shear at each floor level.
2.6.2 Examples: Wind Load Analyses for Buildings #1 and #2
For Building #1, determine the moments, shears, and axial forces using the Portal Method for an interior frame
resulting from wind loads acting in the N-S direction. The wind loads are determined in Section 2.2.1.2.
Moments, shears, and axial forces are shown directly on the frame diagram in Fig. 2-13. The values can be
easily determined by using the following procedure:
(1) Determine the shear forces in the columns:
For the end columns:
3rd story: V = 12.0 kips/6 = 2.0 kips
2nd story V = (12.0 kips + 23.1 kips)/6 = 5.85 kips
1st story: V = (12.0 kips + 23.1 kips + 21.7 kips)/6 = 9.50 kips
The shear forces in the interior columns are twice those in the end columns.
Wh
2n 1
and
Wh
2n n
, n
Wh
2n 1
−
Wh
2n 2
Wh
2n 2
−
Wh
2n 3

2-21
W B C DA
B C D
A
W
˜1/2 ˜2/2 ˜3/2
˜1 ˜2 ˜3
h
h/2
h/2
W
6
W
6
W
6
W
6
W
6
W
6
W
6
W
6
W
6
W
6
W
6
Wh
6˜
Wh
6˜
Wh
6˜
Wh
6˜
Wh
6˜
Wh
6˜
Wh
6˜
Wh
6˜
Wh
6˜
Wh
6˜
Wh
6˜
Wh
6˜
˜1 = ˜2 = ˜3 = ˜
• Assumed inflection point at mid-length members
W
3
W
3
W
3
W
3
W
3
W
3
Wh
6
Wh
6
W
6
Wh
12
Wh
12
Σ = 0 Σ = 0
˜/2˜/2
h/2
B
Wh
12
Wh
12 Wh
12
Wh
6
W
3
Wh
6˜
Wh
6˜
Figure 2-11 Portal Method
Figure 2-12 Joint Detail

2-22
(2) Determine the axial loads in the columns:
For the end columns, the axial loads can be obtained by summing moments about the column
inflection points at each level. For example, for the 2nd story columns:
= 0 : 12(13 + 6.5) + 23.1 (6.5) – P (90) = 0
P = 4.27 kips
For this frame, the axial forces in the interior columns are zero.
(3) Determine the moments in the columns:
The moments can be obtained by multiplying the column shear force by one-half of the column length.
For example, for an exterior column in the 2nd story:
M = 5.85(13/2) = 38.03 ft-kips
(4) Determine the shears and the moments in the beams:
These quantities can be obtained by satisfying equilibrium at each joint. Free-body diagrams for the
2nd story are shown in Fig. 2-14.
As a final check, sum moments about the base of the frame:
= 0: 12.0(39) + 23.1(26) + 21.7(13) – 10.91(90) – 2(61.53 + 123.07) = 0 (checks)
In a similar manner, the wind load analyses for an interior frame of Building #2 (5-story flat plate), in both the
N-S and E-W directions are shown in Figs. 2-15 and 2-16, respectively. The wind loads are determined in
Section 2.2.2.1.
∑M
∑M

Figure 2-13 Shear, Moments and Axial Forces Resulting from Wind Loads for an Interior Frame of
Building #1 in the N-S Direction, using the Portal Method
2-23
Shear forces and axial forces are
in kips, bending moments are in
ft-kips
13'-0"13'-0"13'-0"
12.0 kips M = 13.00 M = 13.00 M = 13.00
V = 0.87 V = 0.87 V = 0.87
30'-0" 30'-0" 30'-0"
23.1 kips M = 51.03 M = 51.03 M = 51.03
V = 3.4V = 3.4V = 3.4
21.7 kips M = 99.56 M = 99.56 M = 99.56
V = 6.64V = 6.64V = 6.64
V = 2.00
M = 13.00
P = 0.87
V = 4.00
M = 26.00
P = 0.00
V = 4.00
M = 26.00
P = 0.00
V = 2.00
M = 13.00
P = 0.87
V = 5.85
M = 38.03
P = 4.27
V = 11.70
M = 76.05
P = 0.00
V = 11.70
M = 76.05
P = 0.00
V = 5.85
M = 38.03
P = 4.27
V = 9.47
M = 61.53
P = 10.91
V = 18.93
M = 123.07
P = 0.00
V = 18.93
M = 123.07
P = 0.00
V = 9.47
M = 61.53
P = 10.91

2-24
0.87k
4.27k
5.85k
2.0k
13.0ft-kips
51.03ft-kips
51.03ft-kips
51.03ft-kips51.03ft-kips
51.03ft-kips
51.03ft-kips
0.87k
4.27k
5.85k
2.0k
13.0ft-kips
23.1k
38.03ft-kips
19.75k
3.4k3.4k3.4k
3.4k3.4k3.4k3.4k3.4k3.4k
3.4k
3.4k
3.4k
3.4k
38.03ft-kips
6.35k
6.35k
26.0ft-kips
4.0k
11.7k
76.05ft-kips
13.05k13.05k
11.7k
76.05ft-kips
26.0ft-kips
4.0k
19.75k
51.03ft-kips
51.03ft-kips
Figure2-14ShearForces,AxialForces,andBendingMomentsat2ndStoryofBuilding#1

Figure 2-15 Shears, Moments, and Axial Forces Resulting from Wind Loads for an Interior Frame of Building
#2 in the N-S Direction, using the Portal Method
2-25
12'-0"12'-0"12'-0"
3.20 kips M = 3.20 M = 3.20 M = 3.20
V = 0.32 V = 0.32 V = 0.32
6.3 kips M = 12.70 M = 12.70 M = 12.70
V = 1.27V = 1.27V = 1.27
6.1 kips M = 25.1 M = 25.1 M = 25.1
V = 2.51V = 2.51V = 2.51
V = 0.53
M = 3.20
P = 0.32
V = 1.07
M = 6.40
P = 0.00
V = 1.07
M = 6.40
P = 0.00
V = 0.53
M = 3.20
P = 0.32
V = 1.58
M = 9.50
P = 1.59
V = 3.17
M = 19.00
P = 0.00
V = 3.17
M = 19.00
P = 0.00
V = 1.58
M = 9.50
P = 1.59
V = 2.60
M = 15.60
P = 4.10
V = 5.20
M = 31.20
P = 0.00
V = 5.20
M = 31.20
P = 0.00
V = 2.60
M = 15.60
P = 4.10
15'-0"12'-0"
5.8 kips M = 37.00 M = 37.00 M = 37.00
V = 3.70V = 3.70V = 3.70
6.1 kips M = 55.78 M = 55.78 M = 55.78
V = 5.58V = 5.58V = 5.58
V = 3.57
M = 21.40
P = 7.80
V = 9.90
M = 59.38
P = 0.00
V = 9.90
M = 59.38
P = 0.00
V = 3.57
M = 21.40
P = 7.80
V = 4.58
M = 34.38
P = 13.38
V = 9.17
M = 68.75
P = 0.00
V = 9.17
M = 68.75
P = 0.00
V = 4.58
M = 34.38
P = 13.38
ft-kips 20'-0" 20'-0" 20'-0"

2-26
Figure 2-16 Shears, Moments, and Axial Forces Resulting from Wind Loads for an Interior Frame of
Building #2 in the E-W Direction, using the Portal Method
References
2.1 American Society of Civil Engineers Minimum Design Loads for Buildings and Other Structures,
ASCE/SEI 7-05, American Society of Civil Engineers, New York, N.Y., 2005.
2.2 Continuity in Concrete Building Frames, Portland Cement Association, Skokie, EB033, 1959, 56 pp.
12'-0"12'-0"12'-0"
2.30 kips M = 1.38 M = 1.38 M = 1.38
V = 0.12 V = 0.12 V = 0.12
4.40 kips M = 5.40 M = 5.40 M = 5.40
V = 0.45V = 0.45V = 0.45
4.20 kips M = 25.1 M = 25.1 M = 25.1
V = 2.51V = 2.51V = 2.51
V = 0.23
M = 1.38
P = 0.12
V = 0.46
M = 2.76
P = 0.00
V = 0.46
M = 2.76
P = 0.00
V = 0.46
M = 2.76
P = 0.00
V = 0.67
M = 4.02
P = 0.57
V = 1.34
M = 8.04
P = 0.00
V = 1.34
M = 8.04
P = 0.00
V = 1.34
M = 8.04
P = 0.00
V = 1.09
M = 6.54
P = 1.45
V = 2.18
M = 13.08
P = 0.00
V = 2.18
M = 13.08
P = 0.00
V = 2.18
M = 13.08
P = 0.00
15'-0"12'-0"
4.00 kips M = 15.48 M = 15.48 M = 15.48
V = 1.29V = 1.29V = 1.29
4.10 kips M = 23.19 M = 23.19 M = 23.19
V = 1.93V = 1.93V = 1.93
V = 1.49
M = 8.94
P = 2.74
V = 2.98
M = 17.88
P = 0.00
V = 2.98
M = 17.88
P = 0.00
V = 2.98
M = 17.88
P = 0.00
V = 1.90
M = 14.25
P = 4.67
V = 3.80
M = 28.50
P = 0.00
V = 3.80
M = 28.50
P = 0.00
V = 3.80
M = 28.50
P = 0.00
ft-kips 24'-0" 24'-0" 24'-0"
M = 1.38 M = 1.38
V = 0.12 V = 0.12
M = 5.40 M = 5.40
V = 0.45V = 0.45
M = 25.1 M = 25.1
V = 2.51V = 2.51
V = 0.46
M = 2.76
P = 0.00
V = 0.23
M = 1.38
P = 0.12
V = 1.34
M = 8.04
P = 0.00
V = 0.67
M = 4.02
P = 0.57
V = 2.18
M = 13.08
P = 0.00
V = 1.09
M = 6.54
P = 1.45
M = 15.48 M = 15.48
V = 1.29V = 1.29
M = 23.19 M = 23.19
V = 1.93V = 1.93
V = 2.98
M = 17.88
P = 0.00
V = 1.49
M = 8.94
P = 2.74
V = 3.80
M = 28.50
P = 0.00
V = 1.90
M = 14.25
P = 4.67
24'-0" 24'-0"

3-1
Chapter 3
Simplified Design for Beams
and One-Way Slabs
3.1 INTRODUCTION
The simplified design approach for proportioning beams and slabs (floor and roof members) is based in part
on published articles,3.1-3.6
and in part on simplified design aid material published by CRSI.3.7,3.10
Additional data
for design simplification are added where necessary to provide the designer with a total simplified design
approach for beam and slab members. The design conditions that need to be considered for proportioning the
beams and slabs are presented in the order generally used in the design process.
The simplified design procedures comply with the ACI 318-11 code requirements for both member strength and
member serviceability. The simplified methods will produce slightly more conservative designs within the limitations
noted.All coefficients are based on the Strength Design Method, using appropriate load factors and strength reduction
factors specified in ACI 318. Where simplified design requires consideration of material strengths, 4000 psi concrete
and Grade 60 reinforcement are used. The designer can easily modify the data for other material strengths.
The following data are valid for reinforced concrete flexural members with › = 4000 psi and fy
= 60,000 psi:
modulus of elasticity for concrete Ec
= 3,600,000 psi (ACI 8.5.1)
modulus of elasticity for rebars Es
= 29,000,000 psi (ACI 8.5.2)
minimum reinforcement ratio (beams, joists) ρmin
= 0.0033 (ACI 10.5.1)
minimum reinforcement ratio (slabs) ρmin
= 0.0018 (ACI 7.12.2) and (ACI 10.5.4)
maximum reinforcement ratio* ρmax
= 0.0206 (ACI 10.3.5)
maximun useful reinforcement ratio** ρt
= 0.0181 (ACI 10.3.4)
3.2 DEPTH SELECTION FOR CONTROL OF DEFLECTIONS
Deflection of beams and one-way slabs need not be computed if the overall member thickness meets the
minimum specified in ACI Table 9.5(a). Table 9.5(a) may be simplified to six values as shown in Table 3-1. The
quantity n is the clear span length for cast-in-place beam and slab construction. For design convenience, min-
imum thicknesses for the six conditions are plotted in Fig. 3-1.
Deflections are not likely to cause problems when overall member thickness meets or exceeds these values for
uniform loads commonly used in the design of buildings. The values are based primarily on experience and are
not intended to apply in special cases where beam or slab spans may be subject to heavily distributed loads or
* The ACI 318-11 does not provide direct maximum reinforcement ratio for beams and slabs. The maximum reinforcement
ratio was derived from the strain profile as limited in ACI Section 10.3.5.
** For tension controlled sections with φ=0.9

concentraded loads. Also, they are not intended to apply to members supporting or attached to nonstructural
elements likely to be damaged by deflections (ACI 9.5.2.1).
Table 3-1 Minimum Thickness for Beams and One-Way Slabs
3-2
Beams and One-way Slabs Minimum h
Simple Span Beams or Joists* /16
Beams or Joists Continuous at one End /18.5
Beams or Joists Continuous at both Ends /21
Simple Span solid Slabs* /20
Solid Slabs Continuous at one End /24
Solid Slabs Continuous at both Ends /28
* Minimum thickness for cantilevers can be considered equal to twice that for a
simple span.
15
20
25
30
35
0
5
10
15
20
25
30
35
5 10 15 20 25 30 35 40 50
MinimumThickness,h(in.)
Clear Span, n (ft)
Simple span
beams or joist
Continuous
beams or joist–
one end cont.
Simple span
slabs
Continuous
slabs–one
end cont.
Continuous beams
or joist–both ends cont.
Continuous slabs
both ends cont.
Figure 3-1 Minimum Thicknessees for Beams and One-Way Slabs

3-3
Chapter 3 • Simplified Design for Beams and Slabs
For roof beams and slabs, the values are intended for roofs subjected to normal snow or construction live loads
only, and with minimal water ponding or drifting snow problems.
Prudent choice of steel percentage can also minimize deflection problems. Members will usually be of
sufficient size, so that deflections will be within acceptable limits, when the tension reinforcement ratio ρ used
in the positive moment regions does not exceed approximately one-half of the maximum value permitted.
For › = 4000 psi and fy
= 60,000 psi, one-half of ρmax
is approximately one percent ( 0.01).
Depth selection for control of deflections of two-way slabs is given in Chapter 4.
As a guide, the effective depth d can be calculated as follows:
For beams with one layer of bars d = h - (ഡ 2.5 in.)
For joists and slabs d = h - (ഡ1.25 in.)
3.3 MEMBER SIZING FOR MOMENT STRENGTH
As noted above, deflection problems are rarely encountered with beams having a reinforcement ratio ρ equal
to about one-half of the maximum permitted. The ACI 318-11 does not provide direct maximum reinforcement
ratio for beams and slabs. The code indirectly defines reinforcements in terms of the net tensile strain in
extreme layer of longitudinal tension steel at nominal strength εt
. For beams, slabs and members with factored
axial compressive load less than 0.10›Ag
, εt
at nominal strength should not be less than 0.004. Figure 3-2
summarizes the code limitations and related definitions for the strain εt
. For our selected materials
(› = 4000 psi and fy
= 60,000 psi), the maximum reinforcement ratio corresponding to εt
= 0.004, is ρmax
=
0.0206. It is important to point out that the increase in the nominal moment capacity Mn
when using larger area
of reinforcement beyond the area associated with the strain limit of tension controlled section (εt
= 0.005), is
offset by the required decrease in the in the φ factor when calculating the design moment capacity φMn
.
To calculate the required dimensions for a member subjected to a factored moment Mu
, nominal flexural
strength equation for the section needs to be developed. The section’s nominal flexural strength can be
calculated based on equilibrium and strain compatibility assuming the strain profile and the concrete stress
block shown in Figure 3-2 (See design assumption of ACI 318-11 Section 10.2). A simplified sizing equation
can be derived using the strength design approach developed in Chapter 6 of Reference 3.8.
Set ρ = 0.5ρmax = 0.0103
M = φAs
fy
(d − a / 2) = φρbdfy
(d − a / 2)
a = As
fy
/ 0.85 ʹfc
b = ρdfy
/ 0.85 ʹfc
Mu
/ φbd2
= ρfy
1−
0.5ρfy
0.85 ʹfc
⎛
⎝
⎜
⎞
⎠
⎟ = Rn
Rn
= ρfy
1−
0.5ρfy
0.85 ʹfc
⎛
⎝
⎜
⎞
⎠
⎟
u

For simplicity, set bd2
reqd
= 20Mu
which corresponds to ρ = 0.0125
= 60,000 psi:
bd2
= 20Mu
where Mu
is in ft-kips and b and d are in inches
A similar sizing equation can be derived for other material strengths. Figure 3-3 shows values for section sizing
for different concrete strengths and reinforcement ratios.
With factored moments Mu
and effective depth d known, the required beam width b is easily determined using
the sizing equation bd2
= 20Mu
(for › = 4000 psi and fy
= 60,000 psi). When frame moments vary, b is
usually determined for the member which has the largest Mu
; for economy, this width may be used for all
similar members in the frame. Since slabs are designed by using a 1-ft strip (b = 12 in.), the sizing equation
can be used to check the initial depth selected for slabs; it simplifies to d = 1.3 .
If the depth determined for control of deflections is shallower than desired, a larger depth may be selected with
a corresponding width b determined from the above sizing equation. Actually, any combination of b and d could
be determined from the sizing equation with the only restriction being that the final depth selected must be
greater than that required for deflection control (Table 3-1).
It is important to note that for minimum beam size with maximum reinforcement, the sizing equation becomes
bd2
min
= 14.6Mu
.
3.3.1 Notes on Member Sizing for Economy
• Use whole inches for overall beam dimensions; slabs may be specified in 1/2
-in. increments.
• Use beam widths in multiples of 2 or 3 inches, such as 10, 12, 14, 16, 18, etc.
• Use constant beam size from span to span and vary reinforcement as required.
• Use wide flat beams (same depth as joist system) rather than narrow deep beams.
• Use beam width equal to or greater than the column width.
• Use uniform width and depth of beams throughout the building.
See also Chapter 9 for design considerations for economical formwork.
Mu
3-4
= 0.0103× 60,000 1−
0.5 × 0.0103× 60,000
0.85 × 4000
⎡
⎣⎢
⎤
⎦⎥
= 562psi
bd2
reqd =
Mu
Rn
=
Mu
×12 ×1000
0.9 × 562
= 23.7Mu
φ

3-5
εt ≥ 0.005 Tension control φ = 0.9
εt = 0.004 Minimum for beams and slabs φ = 0.817
εt = 0.002 Balanced condition φ = 0.65 (See chapter 5 Columns)
εt ≤ 0.002 Compression controls φ = 0.65 (See chapter 5 Columns)
0.005 > εt < 0.002 Transition
C = 0.85f’c
T = FsAs ≤ FyAs
T = C
b Strain
d
c
0.003
T
C
εt
a=β1c
0.85f’c
Stress
Mn = C [d –a/2] = T (d-a/2)
d - —a
2
Figure 3-2 Code Strain Distribution and Stress Block for Nominal Strength Calculations
3.4 DESIGN FOR FLEXURAL REINFORCEMENT
Similar to developing the sizing equation, a simplified equation for the area of tension steel As
can be derived
using the strength design approach developed in Chapter 6 of Reference 3.8. An approximate linear relationship
between Rn
and ρ can be described by an equation in the form Mn
/bd2
= ρ (constant), which readily converts to
As
= Mu
/φd(constant). This linear equation for As
is reasonably accurate up to about two-thirds of the maximum
ρ. For › = 4000 psi and ρ = 0.0125 (60% ρmax
), the constant for the linear approximation is :

3-6
*To convert Mu
from ft-kips to in.-lbs
Therefore,
= 60,000 psi:
**Note: this equation is in mixed units: Mu
is in ft-kips, d is in in. and As
is in sq in.
For all values of ρ < 0.0125, the simplified As
equation is slightly conservative. The maximum deviation in
As
is less than + 10% at the minimum and maximum useful tension steel ratios.3.9
For members with
reinforcement ratios in the range of approximately 1% to 1.5%, the error is less than 3%.
The range of change in the constant in the denominator in the above equation is very narrow for different
concrete strengths; this allows the use of the above equation for approximate reinforcement area estimination
with other concrete strengths.
The simplified As
equation is applicable for rectangular cross sections with tension reinforcement only.
Members proportioned with reinforcement in the range of 1% to 1.5% will be well within the code limits for
singly reinforced members. For positive moment reinforcement in flanged floor beams, As
is usually computed
for a rectangular compression zone; rarely will As
be computed for a T-shaped compression zone.
The depth of the rectangular compression zone, a, is given by:
where be
= effective width of slab as a T-beam flange (ACI 8.12).
The flexural member is designed as a rectangular section whenever hf ≥ a where hf is the thickness of the slab
(i.e., flange thickness).
For more accurate calculations, Figure 3-4 shows values for the constant required to calculate the reinforcement
for different concrete strengths and reinforcement ratio. It is important to note that Figures 3-3 and 3-4 can
be used for sizing and reinforcement calculations for rectangular sections for any concrete strength and
reinforcement ratio.
3.5 REINFORCING BAR DETAILS
The minimum and maximum number of reinforcing bars permitted in a given cross section is a function of
cover and spacing requirements given in ACI 7.6.1 and ACI 3.3.2 (minimum spacing for concrete placement),
ACI 7.7.1 (specified cover for protection of reinforcement), and ACI 10.6 (maximum spacing for control of
As =
Mu
**
4d
a =
Asfy
0.85 ʹfc
be
fy
12,000 *
1−
0.5ρfy
0.85 ʹfc
⎡
⎣
⎢
⎤
⎦
⎥ =
60,000
12,000
1−
0.5(0.0125)(60)
0.85 × 4
⎡
⎣⎢
⎤
⎦⎥ = 4.45
As
=
Mu
φd(constant)
=
Mu
0.9 × 4.45 × d
=
Mu
4d

3-7
0.0075 0.01 0.0125 0.015 0.0175 0.02
32
28
24
20
16
12
Constant(1)
bd2 = (Constant(1))Mu
3 ksi
4 ksi
5 ksi
6 ksi
7 ksi
8 ksi
ρ
0.0075 0.01 0.0125 0.015 0.0175 0.02
4.40
4.20
4.00
3.80
3.60
3.40
Constant(2)
As =
3 ksi
4 ksi
5 ksi
6 ksi
7 ksi
8 ksi
ρ
(Constant(2))d
Mu
Figure 3-3 Constant for Section Sizing
Figure 3-4 Constant for Area of Reinforcement Calculations

3-8
flexural cracking). Tables 3-2 and 3-3 give the minimum and maximum number of bars in a single layer for
beams of various widths; selection of bars within these limits will provide automatic code conformance with
the cover and spacing requirements.
Table 3-2 Minimum Number of Bars in a Single Layer (ACI 10.6)
The values in Tables 3-2 are based on a cover of 2 in. to the main flexural reinforcement (i.e., 1.5 in. clear
cover to the stirrups plus the diameter of a No. 4 stirrup). In general, the following equations can be used to
determine the minimum number of bars n in a single layer for any situation (see Fig. 3-2):
where
Table 3-3 Maximum Number of Bars in a Single Layer
nmin =
bw
− 2(cc
+ 0.5db
)
s
+1
s = 15
40,000
fs
⎛
⎝
⎜
⎞
⎠
⎟ − 2.5cc
≤ 12
4,000
fs
⎛
⎝
⎜
⎞
⎠
⎟
BAR BEAM WIDTH (in.)
SIZE 12 14 16 18 20 22 24 26 28 30 36 42 48
No. 4 5 6 8 9 10 12 13 14 16 17 21 25 29
No. 5 5 6 7 8 10 11 12 13 15 16 19 23 27
No. 6 4 6 7 8 9 10 11 12 14 15 18 22 25
No. 7 4 5 6 7 8 9 10 11 12 13 17 20 23
No. 8 4 5 6 7 8 9 10 11 12 13 16 19 22
No. 9 3 4 5 6 7 8 8 9 10 11 14 17 19
No. 10 3 4 4 5 6 7 8 8 9 10 12 15 17
No. 11 3 3 4 5 5 6 7 8 8 9 11 13 15
BAR BEAM WIDTH (in.)
SIZE 12 14 16 18 20 22 24 26 28 30 36 42 48
No. 4 2 2 3 3 3 3 3 4 4 4 5 5 6
No. 5 2 2 3 3 3 3 3 4 4 4 5 5 6
No. 6 2 2 3 3 3 3 3 4 4 4 5 5 6
No. 7 2 2 3 3 3 3 3 4 4 4 5 5 6
No. 8 2 2 3 3 3 3 3 4 4 4 5 5 6
No. 9 2 2 3 3 3 3 3 4 4 4 5 5 6
No. 10 2 2 3 3 3 3 3 4 4 4 5 5 6
No. 11 2 2 3 3 3 3 3 4 4 4 5 5 6

3-9
Figure 3-5 Cover and Spacing Requirements for Tables 3-2 and 3-3
where
bw
= beam width, in.
cc
= clear cover to tension reinforcement, in.
cs
= clear cover to stirrups, in.
db
= diameter of main flexural bar, in.
ds
= diameter of stirrups
The values obtained from the above equations should be rounded up to the next whole number.
The values in Table 3-3 can be determined from the following equation:
where
The minimum clear space between bars is defined in Fig. 3-5. The above equation can be used to determine the
maximum number of bars in any general case; computed values should be rounded down to the next whole number.
Suggested temperature and shrinkage reinforcement for one-way floor and roof slabs is given in Table 3-4.
The provided area of reinforcement (per foot width of slab) satisfies ACI 7.12.2. Bar spacing must not exceed
5h and 18 in. (where h = thickness of slab). The same area of reinforcement is also applied for minimum
moment reinforcement in one-way slabs (ACI 10.5.4) at a maximum spacing of 3h, not to exceed 18 in.
(ACI 7.6.5). As noted in Chapter 4, this same minimum area of steel applies for flexural reinforcement in
each direction for two-way floor and roof slabs; in this case, the maximum spacing is 2h not to exceed 18 in.
(ACI 13.3).
As an aid to designers, reinforcing bar data are presented in Tables 3-5 and 3-6.
See Chapter 8, Section 8.2, for notes on reinforcement selection and placement for economy.
n = 1+
(minimum clear space) + db
bw
− 2(cs
+ ds
+ r)
r =
3/4in. for No. 3 stirrups
1in. for No. 4 stirrups
⎧
⎨
⎩
cs ds
db
r
#3 stirrup with #5 &
#6 bars
#4 stirrup with
#7—#11 bars
cs
11/2" min.
clear to
stirrups
(ACI 7.7.1)
r =
3/4" for #3 stirrups
1" for #4 stirrups
Minimum
clear space
= largest of
1"
db
(1.33) (max. aggregate size)

Table 3-4 Temperature Reinforcement for One-Way Slabs
3-10
Number of bars
Bar
size
Bar diameter
(in.) 1 2 3 4 5 6 7 8
#3 0.375 0.11 0.22 0.33 0.44 0.55 0.66 0.77 0.88
#4 0.500 0.20 0.40 0.60 0.80 1.00 1.20 1.40 1.60
#5 0.625 0.31 0.62 0.93 1.24 1.55 1.86 2.17 2.48
#6 0.750 0.44 0.88 1.32 1.76 2.20 2.64 3.08 3.52
#7 0.875 0.60 1.20 1.80 2.40 3.00 3.60 4.20 4.80
#8 1.000 0.79 1.58 2.37 3.16 3.95 4.74 5.53 6.32
#9 1.128 1.00 2.00 3.00 4.00 5.00 6.00 7.00 8.00
#10 1.270 1.27 2.54 3.81 5.08 6.35 7.62 8.89 10.16
#11 1.410 1.56 3.12 4.68 6.24 7.80 9.36 10.92 12.48
Table 3-6 Areas of Bars per Foot Width of Slab— As (in.2
/ft)
Bar Bar spacing (in.)
size 6 7 8 9 10 11 12 13 14 15 16 17 18
#3 0.22 0.19 0.17 0.15 0.13 0.12 0.11 0.10 0.09 0.09 0.08 0.08 0.07
#4 0.40 0.34 0.30 0.27 0.24 0.22 0.20 0.18 .017 0.16 0.15 0.14 0.13
#5 0.62 0.53 0.46 0.41 0.37 0.34 0.31 0.29 0.27 0.25 0.23 0.22 0.21
#6 0.88 0.75 0.66 0.59 0.53 0.48 0.44 0.41 0.38 0.35 0.33 0.31 0.29
#7 1.20 1.03 0.90 0.80 0.72 0.65 0.60 0.55 0.51 0.48 0.45 0.42 0.40
#8 1.58 1.35 1.18 1.05 0.95 0.86 0.79 0.73 0.68 0.63 0.59 0.56 0.53
#9 2.00 1.71 1.50 1.33 1.20 1.09 1.00 0.92 0.86 0.80 0.75 0.71 0.67
#10 2.54 2.18 1.91 1.69 1.52 1.39 1.27 1.17 1.09 1.02 0.95 0.90 0.85
#11 3.12 2.67 2.34 2.08 1.87 1.70 1.56 1.44 1.34 1.25 1.17 1.10 1.04
Table 3-5 Total Areas of Bars—As
(in.2
)
Table 3-6 Total Area of Bars per Foot Width of Slab—As
(in.2
/ft)
p y
Slab Thickness
h (in.)
As (req’d)*
(in.2
/ft)
Suggested
Reinforcement**
3-½ 0.08 #3@16
4 0.09 #3@15
4-½ 0.10 #3@13
5 0.11 #3@12
5-½ 0.12 #4@18
6 0.13 #4@18
6-½ 0.14 #4@17
7 0.15 #4@16
7-½ 0.16 #4@15
8 0.17 #4@14
8-½ 0.18 #4@13
9 0.19 #4@12
9-½ 0.21 #5@18
10 0.22 #5@17
*As = 0.0018bh = 0.022h (ACI 7.12.2).
**For minimum moment reinforcement, bar spacing must not exceed 3h
or 18 in. (ACI 7.6.5). For 3½ in. slab, use #3@10 in.; for 4 in. slab, use
#3@12 in.; for 5½ in. slab, use #3@11 in. or #4 @16 in.

3-11
3.6 DESIGN FOR SHEAR REINFORCEMENT
In accordance with ACI Eq. (11-2), the total shear strength is the sum of two components: shear strength
provided by concrete (φVc) and shear strength provided by shear reinforcement (φVs). Thus, at any section of
the member, Vu ≤ φVc + φVs. Using the simplest of the code equations for shear strength of concrete, Vc, can
be calculated as follows:
Vc = φ2λ bwd ACI Eq. (11−3)
where λ is a modification factor to account for the reduced mechanical properties of lightweight concrete.
λ = 1.0 for normal weigh concrete, λ = 0.85 for sand-lightweight concrete and 0.75 for all-lightweight
concrete. Figure 3-6 shows the values for nominal shear strength provided by concrete (Vc) for a rectangular
section, for normal and light weight concrete and for different concrete strengths. Figure 3-7 and Table 3-7
summarize ACI 318 provisions for shear reinforcement.
ʹfc
0.08
0.1
0.12
0.14
0.16
0.18
3 4 5 6 7 8
ConcreteShearStrengthvc(ksi)
f'c (ksi)
Vc = φ vcbwd
Normal weight
Sand lightweight
All lightweight
Figure 3-6 Nominal Shear Strength Provided by Concrete Vc (Kips)

3-12
Table 3-7 ACI Provisions for Shear Design
* Members subjected to shear and flexure only; φVc=φ2 bwd, φ= 0.75 (ACI 11.2.1.1)
** Av
= 2 ϫ Ab
for U stirrups; fy
≤ 60 ksi (ACI 11.4.2)
*** Maximum spacing based on minimum shear reinforcement (= Av
fy
/50bw
) must also be considered (ACI 11.4.6.3).
* Vs ≤ 8 bwd
** See ACI Section 11.4.6.1 for exceptions
Vu max
φVc
0.5φVc
d/2
Zone 1 Zone 2 Zone 3
φVc ≥ Vu > 0.5 φVcVu > Vc Vu≤ 0.5 φVc
Provide calculated shear
reinforcement (φVs = Vu – φVc)*
Provide minimum shear
reinforcement**
No shear
reinforcement
required
φ
Figure 3-7 Shear Reinforcement Requirements
ʹfc

3-13
The design values in table 3-8 are valid for › = 4000 psi:
The selection and spacing of stirrups can be simplified if the spacing is expressed as a function of the effective depth
d (see Reference 3.3). According to ACI 11.4.5.1 and ACI 11.4.5.3, the practical limits of stirrup spacing vary
from s = d/2 to s = d/4, since spacing closer than d/4 is not economical. With one intermediate spacing at d/3, the
calculation and selection of stirrup spacing is greatly simplified. Using the three standard stirrup spacings noted
above (d/2, d/3, and d/4), a specific value of φVs
can be derived for each stirrup size and spacing as follows:
For vertical stirrups:
ACI Eq. (11.15)
By substituting d/n for s (where n = 2, 3, or 4), the above equation can be rewritten as:
Thus, for No.3 U-stirrups @ s = d/2 with fy
= 60,000 psi and φ = 0.75
φVs
= 0.75(0.22)60 ϫ 2 = 19.8 kips, say 20 kips
The values φVs
given in Table 3-9 may be used to select shear reinforcement with Grade 60 rebars.
Table 3-9 Values of φVs
(fy
= 60 ksi)
It should be noted that these values of φVs are not dependent on the member size nor on the concrete strength.
In the above equations, bw
and d are in inches and the resulting shears are in kips.
s #3 U-stirrups #4 U-stirrups #5 U-stirrups
d/2 20 kips 36 kips 56 kips
*Valid for stirrups with 2 legs (double the tabulated values for 4 legs, etc.)
φVs
=
φAv
fy
d
s
φVs
= φAv
fy
n
Equation Design Value ACI Section
Vc = 2 bwd 0.095bwd ACI 11.2.1.1
0.50 Vc = bwd 0.048bwd ACI 11.4.6.1
Maximum ( Vc + Vs) = 10 bwd 0.48bwd ACI 11.4.7.9
Joists defined by ACI 8.13
Vc = 2.2 bwd
0.104bwd ACI 8.13.8
* bw and d are in inches and the resulting shear in kips
ʹfc
ʹfc
ʹfc
ʹfc
Table 3-8 Concrete Shear Strength Design Values for › = 4000 psi

3-14
The design charts in Figs. 3-8 through 3-11 offer another simplified method for shear design. By entering the charts
with values of d and φVs
= Vu
- φVc
for the member at the section under consideration, the required stirrup spacing
can be obtained by locating the first line above the point of intersection of d and φVs
. Values for spacing not shown
can be interpolated from the charts if desired. Also given in the charts the values for the minimum practical beam
widths bw
that correspond to the maximum allowable for each given spacing s; any member
which has at least this minimum bw
will be adequate to carry the maximum applied Vu
. Fig. 3-11 can also be used
to quickly determine if the dimensions of a given section are adequate: any member with an applied Vu
which is
less than the applicable Vu(max)
can carry this shear without having to increase the values of bw
and/or d. Once the
adequacy of the cross-section has been verified, the stirrup spacing can be established by using Figs. 3-8 through
3-11. This spacing must then be checked for compliance with all maximum spacing criteria.
φVs
= φ8 ʹfc
bw
d

Figure 3-8 Design Chart for Stirrup Spacing, #3-U Stirrups
3-15
10
20
30
40
50
60
70
80
10 15 20 25 30 35 40
d, in
18"
16"
14"
12"
10"
8"
6"
spacing, s = 4"
Vu − φVc =
φAvfyd
s
φVs=Vu−φVc,kips
* Horizontal line indicates for s = d/2.
** Minimum bw
corresponding to bw
d is less than 8 in. for all s.
φVs
φVs = φ8 ʹfc

3-16
30
40
50
60
70
80
90
10 15 20 25 30 35 40
d, in
φVs=Vu−φVc,kips
spacing, s = 4"
Vu − φVc =
φAvfyd
s
18"
(—)
(—)
(—)
(—)
(—)
(—)
(8")
(12")**
16"
14"
12"
10"
8"
6"
** Values in ( ) indicate minimum practical bw
corresponding to bw
d for given s.
(–) Indicates minimum bw
corresponding to bw
d is less than 8 in. for given s.
φVs
φVs = φ8 ʹfc
φVs = φ8 ʹfc

3-17
50
60
70
80
90
100
110
120
10 15 20 25 30 35 40
d, in
spacing,
s = 4"
Vu − φVc =
φAvfyd
s
(—)
(—)
(—)
(—)
(20")**
6" (14")
8"
(10")
10"
(8")
12"
14"
16"
18"
φVs=Vu−φVc,kips
** Values in ( ) indicate minimum practical bw
corresponding to bw
d for given s.
(–) Indicates minimum bw
corresponding to bw
d is less than 8 in. for given s.
φVs
φVs = φ8 ʹfc
φVs = φ8 ʹfc

3-18
Figure 3-11 - Design Chart for Stirrups Spacing for Different Stirrup Sizes
#5
#4
#3
0.25 0.3 0.35 0.4 0.45 0.5
s/d
φVs=Vu-φVc’kips
130
120
110
100
90
80
70
60
50
40
30
20
10
0

3-19
Figure 3-12 Design Chart for Maximum Shear Force—› = 4 ksi
40
80
120
160
200
240
280
320
360
400
10 15 20 25 30 35 40
d, in.
bw=36"
30"
24"
22"
20"
18"
16"
14"
12"
10"

3-20
3.6.1 Example: Design for Shear Reinforcement
The example shown in Fig. 3-13 illustrates the simple procedure for selecting stirrups using design values for
Vc and Vs.
(1) Design data: › = 4000 psi, fy = 60,000 psi, wu = 7 kips/ft.
(2) Calculations: Vu @ column centerline: wu˜/2 = 7 ϫ 24/2 = 84.0 kips
Vu @ face of support: 84 – 1.17(7) = 75.8 kips
Vu @ d from support face (critical section): 75.8 – 2(7) = 61.8 kips
(φVc + φVs)max: 0.48 bwd = 0.48(12)(24) = 138.2 kips
φVc: 0.095 bwd= 0.095(12)(24) = 27.4 kips
φVc /2: 0.048 bwd = 0.048(12)(24) = 13.80 kips
(3) Beam size is adequate for shear strength, since 138.2 kips > 61.8 kips (also see Fig. 3-12). φVs
(required)
= 61.8 – 27.4 = 34.4 kips. From Table 3-9, No.4 @ d/2 = 12 in. is adequate for full length where stirrups
are required since φVs
= 36 kips > 34.4 kips. Length over which stirrups are required is (75.8 – 13.8)/7
= 8.86 ft from support face.
Check maximum stirrup spacing:
Since φVs
= 36 kips < 54.6 kips, the maximum spacing is the least of the following:
Use 10-No.4 U-stirrups at 12 in. at each end of beam.
The problem may also be solved graphically as shown in Fig. 3-13. φVs
for No.3 stirrups at d/2, d/3, and d/4
are scaled vertically from φVc. The horizontal intersection of the φVs values (20 kips, 30 kips, and 40 kips) with
the shear diagram automatically sets the distances where the No.3 stirrups should be spaced at d/2, d/3, and d/4.
The exact numerical values for these horizontal distances are calculated as follows (although scaling from the
sketch is close enough for practical design):
No.3 @ d/4 = 6 in.: (75.8 – 57.4)/7 = 2.63 ft (31.5 in.) use 6 @ 6 in.
@ d/3 = 8 in.: (57.4 – 47.4)/7 = 1.43 ft (17.0 in.) use 2 @ 8 in.
@ d/2 = 12 in.: (47.4 – 13.8)/7 = 4.8 ft (57.6 in.) use 5 @ 12 in.
A more practical solution may be to eliminate the 2 @ 8 in. and use 9 @ 6 in. and 5 @ 12 in.
smax
=
d / 2 = 12in.(governs)
24in.
Avfy / 50bw = 40in.
⎧
⎨
⎪⎪
⎩
⎪
⎪
φ4 ʹfc
bw
d = 0.19bw
d = 54.6 kips

3-21
As an alternative, determine the required spacing of the No.3 U-stirrups at the critical section using Fig. 3-8.
Enter the chart with d = 24 in. and φVs
= 38.9 kips. The point representing this combination is shown in the
design chart. The line immediately above this point corresponds to a spacing of s = 6 in. which is exactly what
was obtained using the previous simplified method.
Figure 3-13 Simplified Method for Stirrup Spacing (Example 3.6.1)
CL Span
5 @ 12"2 @ 8"6 @ 6"2"14"
12'-0"
wu = 7 kips/ftCL Column
bw = 12"
d = 24"h = 27"
#3 @ d/2 d/3 d/4
d from support face
Face of
column
84 kips
75.8 kips
slope = 7 kips/ft
61.8 kips
1.17' 2.0'
57.4 kips (27.4 + 30)
47.4 kips ( 27.4 + 20)
27.4 kips = φVc
20 kips
30 kips
40 kips
φVc/2 13.8 kips
Stirrups Required
No
stirrups
Required
1.17'
s = d/4 s = d/3 s = d/2
2.63 1.43 4.8
φVc + φVs = 27.4 + 40 = 67.4 > 61.8 kips
φVs

3.6.2 Selection of Stirrups for Economy
Selection of stirrup size and spacing for overall cost savings requires consideration of both design time and
fabrication and placing costs. An exact solution with an intricate stirrup layout closely following the variation
in the shear diagram is not a cost-effective solution. Design time is more cost-effective when a quick, more
conservative analysis is utilized. Small stirrup sizes at close spacings require disproportionately high costs in
labor for fabrication and placement. Minimum cost solutions for simple placing should be limited to three
spacings: the first stirrup located at 2 in. from the face of the support (as a minimum clearance), an
intermediate spacing, and finally, a maximum spacing at the code limit of d/2. Larger size stirrups at wider
spacings are more cost-effective (e.g., using No.4 for No.3 at double spacing, and No.5 and No.4 at 1.5 spacing)
if it is possible to use them within the spacing limitations of d/2 and d/4.
In order to adequately develop the stirrups, the following requirements must all be satisfied (ACI 12.13):
(1) stirrups shall be carried as close to the compression and tension surfaces of the member as cover
requirements permit, (2) for No.5 stirrups and smaller, a standard stirrup hook (as defined in ACI 7.1.3) shall
be provided around longitudinal reinforcement, and (3) each bend in the continuous portion of the stirrup must
enclose a longitudinal bar. To allow for bend radii at corners of U stirrups, the minimum beam widths given in
Table 3-10 should be provided.
Table 3-10 Minimum Beam Widths for Stirrups
Note that either the No.3 or the No.4 stirrup in the example of Fig. 3-8 can be placed in the 12 in. wide beam.
3.7 DESIGN FOR TORSION
For simplified torsion design for spandrel beams, where the torsional loading is from an integral slab, two
options are possible:
(1) Size the spandrel beams so that torsion effects can be neglected (ACI 11.5.1), or
(2) Provide torsion reinforcement for a prescribed torque corresponding to the cracking torsional
moment (ACI 11.5.2.2)
Stirrup size
Minimum beam width
(bw)
#3 10 in.
#4 12 in.
#5 14 in.
3-22

3-23
3.7.1 Beam Sizing to Neglect Torsion
Torsion can be neglected if the factored torque Tu (threshold torsion) is less than (ACI 11.5.1(a))
where:
Acp = area enclosed by outside perimeter of concrete-cross section, in2
.
pcp = outside perimeter of concrete cross-section, in.
φ = 0.75
λ = modification factor for lightweight concrete
For normal weight concrete with compressive strength = 4000 psi, torsion can be neglected if:
where Acp
and pcp
are in in.2
and in., respectively, and Tu
in ft-kips.
Section properties Acp
and pcp
are presented in Table 3-11 for rectangular and L shaped beams. Table 3-12
presents values for threshold torsion for › = 4 ksi
For a rectangular sections having width and height equal to b and h, respectively, the torsion can be neglected if:
for concrete with › = 4000 psi:
where h and b are in inches, and Tu ft-kip
Figure 3-14 shows values for threshold torsion for rectangular sections.
Tu < 0.004
A2
cp
pcp
⎛
⎝
⎜
⎞
⎠
⎟
Tu
< φ ʹfc
h2
b2
2(h + b)
⎡
⎣
⎢
⎤
⎦
⎥
Tu
< 0.002
h2
b
2
h + b
⎡
⎣⎢
⎤
⎦⎥
φλ ʹfc
A2
cp
pcp
⎛
⎝
⎜
⎞
⎠
⎟

3-24
h = 48”
h = 44”
h = 40”
h = 36”
h = 32”
h = 28”
h = 24”
h = 20”
h = 16”
h = 12”
12 16 20 24 28 32 36 40 44 48
b(in.)
ThresholdTu(ft-kips)
1000
900
800
700
600
500
400
300
200
100
0
b
h
Figure 3-14 Dimensions of Rectangular Cross Section to Neglect Torsion (f'c = 4 ksi)

3-25
Acp pcp Acp pcp Acp pcp Acp pcp Acp pcp Acp pcp Acp pcp
0 12 0 144 48 192 56 240 64 288 72 336 80 384 88 432 96
Rectangular 16 0 192 56 256 64 320 72 384 80 448 88 512 96 576 104
Section 20 0 240 64 320 72 400 80 480 88 560 96 640 104 720 112
24 0 288 72 384 80 480 88 576 96 672 104 768 112 864 120
28 0 336 80 448 88 560 96 672 104 784 112 896 120 1008 128
3212 16 20 24 28 36
slab
thickness
hf (in.)
Beam
depth
h (in.)
Max.
Slab
width
Beam width h (in.)
28 0 336 80 448 88 560 96 672 104 784 112 896 120 1008 128
32 0 384 88 512 96 640 104 768 112 896 120 1024 128 1152 136
36 0 432 96 576 104 720 112 864 120 1008 128 1152 136 1296 144
40 0 480 104 640 112 800 120 960 128 1120 136 1280 144 1440 152
44 0 528 112 704 120 880 128 1056 136 1232 144 1408 152 1584 160
48 0 576 120 768 128 960 136 1152 144 1344 152 1536 160 1728 168
5 12 7 179 62 227 70 275 78 323 86 371 94 419 102 467 110
16 11 247 78 311 86 375 94 439 102 503 110 567 118 631 126
20 15 315 94 395 102 475 110 555 118 635 126 715 134 795 142
24 19 383 110 479 118 575 126 671 134 767 142 863 150 959 158
28 20 436 120 548 128 660 136 772 144 884 152 996 160 1108 168
32 20 484 128 612 136 740 144 868 152 996 160 1124 168 1252 176
36 20 532 136 676 144 820 152 964 160 1108 168 1252 176 1396 184
40 20 580 144 740 152 900 160 1060 168 1220 176 1380 184 1540 192
44 20 628 152 804 160 980 168 1156 176 1332 184 1508 192 1684 200
48 20 676 160 868 168 1060 176 1252 184 1444 192 1636 200 1828 20848 20 676 160 868 168 1060 176 1252 184 1444 192 1636 200 1828 208
6 12 6 180 60 228 68 276 76 324 84 372 92 420 100 468 108
16 10 252 76 316 84 380 92 444 100 508 108 572 116 636 124
20 14 324 92 404 100 484 108 564 116 644 124 724 132 804 140
24 18 396 108 492 116 588 124 684 132 780 140 876 148 972 156
28 22 468 124 580 132 692 140 804 148 916 156 1028 164 1140 172
32 24 528 136 656 144 784 152 912 160 1040 168 1168 176 1296 184
36 24 576 144 720 152 864 160 1008 168 1152 176 1296 184 1440 192
40 24 624 152 784 160 944 168 1104 176 1264 184 1424 192 1584 200
44 24 672 160 848 168 1024 176 1200 184 1376 192 1552 200 1728 208
48 24 720 168 912 176 1104 184 1296 192 1488 200 1680 208 1872 216
Table 3-11 Torsional Section Properties
bf
bf
bf
b
h
4hf
(h-h )f
ACI 13.2.4}
hf
≤
≤

3-26
Acp pcp Acp pcp Acp pcp Acp pcp Acp pcp Acp pcp Acp pcp
7 12 5 179 58 227 66 275 74 323 82 371 90 419 98 467 106
16 9 255 74 319 82 383 90 447 98 511 106 575 114 639 122
20 13 331 90 411 98 491 106 571 114 651 122 731 130 811 138
32 36
slab
thickness
hf (in.)
Beam
depth
h (in.)
Max.
Slab
width
Beam width h (in.)
12 16 20 24 28
20 13 331 90 411 98 491 106 571 114 651 122 731 130 811 138
24 17 407 106 503 114 599 122 695 130 791 138 887 146 983 154
28 21 483 122 595 130 707 138 819 146 931 154 1043 162 1155 170
32 25 559 138 687 146 815 154 943 162 1071 170 1199 178 1327 186
36 28 628 152 772 160 916 168 1060 176 1204 184 1348 192 1492 200
40 28 676 160 836 168 996 176 1156 184 1316 192 1476 200 1636 208
44 28 724 168 900 176 1076 184 1252 192 1428 200 1604 208 1780 216
48 28 772 176 964 184 1156 192 1348 200 1540 208 1732 216 1924 224
8 12 4 176 56 224 64 272 72 320 80 368 88 416 96 464 104
16 8 256 72 320 80 384 88 448 96 512 104 576 112 640 120
20 12 336 88 416 96 496 104 576 112 656 120 736 128 816 136
24 16 416 104 512 112 608 120 704 128 800 136 896 144 992 152
28 20 496 120 608 128 720 136 832 144 944 152 1056 160 1168 168
32 24 576 136 704 144 832 152 960 160 1088 168 1216 176 1344 184
36 28 656 152 800 160 944 168 1088 176 1232 184 1376 192 1520 200
40 32 736 168 896 176 1056 184 1216 192 1376 200 1536 208 1696 21640 32 736 168 896 176 1056 184 1216 192 1376 200 1536 208 1696 216
44 32 784 176 960 184 1136 192 1312 200 1488 208 1664 216 1840 224
48 32 832 184 1024 192 1216 200 1408 208 1600 216 1792 224 1984 232
9 12 3 171 54 219 62 267 70 315 78 363 86 411 94 459 102
16 7 255 70 319 78 383 86 447 94 511 102 575 110 639 118
20 11 339 86 419 94 499 102 579 110 659 118 739 126 819 134
24 15 423 102 519 110 615 118 711 126 807 134 903 142 999 150
28 19 507 118 619 126 731 134 843 142 955 150 1067 158 1179 166
32 23 591 134 719 142 847 150 975 158 1103 166 1231 174 1359 18232 23 591 134 719 142 847 150 975 158 1103 166 1231 174 1359 182
36 27 675 150 819 158 963 166 1107 174 1251 182 1395 190 1539 198
40 31 759 166 919 174 1079 182 1239 190 1399 198 1559 206 1719 214
44 35 843 182 1019 190 1195 198 1371 206 1547 214 1723 222 1899 230
48 36 900 192 1092 200 1284 208 1476 216 1668 224 1860 232 2052 240
10 12 2 164 52 212 60 260 68 308 76 356 84 404 92 452 100
16 6 252 68 316 76 380 84 444 92 508 100 572 108 636 116
20 10 340 84 420 92 500 100 580 108 660 116 740 124 820 13220 10 340 84 420 92 500 100 580 108 660 116 740 124 820 132
24 14 428 100 524 108 620 116 716 124 812 132 908 140 1004 148
28 18 516 116 628 124 740 132 852 140 964 148 1076 156 1188 164
32 22 604 132 732 140 860 148 988 156 1116 164 1244 172 1372 180
36 26 692 148 836 156 980 164 1124 172 1268 180 1412 188 1556 196
40 30 780 164 940 172 1100 180 1260 188 1420 196 1580 204 1740 212
44 34 868 180 1044 188 1220 196 1396 204 1572 212 1748 220 1924 228
48 38 956 196 1148 204 1340 212 1532 220 1724 228 1916 236 2108 244
Table 3-11 Torsional Section Properties (Continued)

3-27
Beam with flange < Beam without flange
Acp
2
/pcp Acp
2
/pcp Acp
2
/pcp Acp
2
/pcp Acp
2
/pcp Acp
2
/pcp Acp
2
/pcp
0 12 0 432 658 900 1152 1411 1676 1944
16 0 658 1024 1422 1843 2281 2731 3190
20 0 900 1422 2000 2618 3267 3938 4629
24 0 1152 1843 2618 3456 4342 5266 6221
28 0 1411 2281 3267 4342 5488 6690 7938
32 0 1676 2731 3938 5266 6690 8192 9758
36 0 1944 3190 4629 6221 7938 9758 11664
40 0 2215 3657 5333 7200 9224 11378 13642
44 0 2489 4130 6050 8200 10540 13043 15682
48 0 2765 4608 6776 9216 11884 14746 17774
5 12 12 432 658 900 1152 1411 1676 1944
16 16 658 1024 1422 1843 2281 2724 3160
20 20 900 1422 2000 2610 3200 3815 4451
24 24 1152 1843 2618 3360 4143 4965 5821
28 28 1411 2281 3203 4139 5141 6200 7308
32 32 1676 2731 3803 4957 6200 7520 8906
17
26
36
46
56
66
26
40
56
73
98 163 239 324
90
108
36
56
79
103
127
150
417 516 620
109 182 268 364 470 583 703
77 126 183 246 314 386 461
88 145 211 285 365 450 539
56 90 129 172 217 264 314
66 108 156 208 264 324 386
36 56 79 103 129 156 183
46 73 103 137 172 208 246
17 26 36 46 56 66 77
26 40 56 73 90 108 126
slab
thickness
hf (in.)
Beam
depth h
(in.)
Max.
Slab
width
bf (in.)
Beam width h (in.)
12 16 20 24 28 32 36
Tu
ft kips
Tu
ft kips
Tu
ft kips
Tu
ft kips
Tu
ft kips
Tu
ft kips
Tu
ft kips
56
90
126
164
203
245
46
73
103
133
164
196
77
125
176
230
289
352
66
108
151
196
245
297
)
p
A
(fT
cp
2
cp'
cu
cp
2
cp
p
A
cp
2
cp
p
A
32 32 1676 2731 3803 4957 6200 7520 8906
36 36 1944 3173 4424 5808 7308 8906 10591
40 40 2215 3603 5063 6688 8457 10350 12352
44 44 2489 4040 5717 7593 9643 11844 14179
48 48 2765 4485 6384 8519 10860 13382 16065
6 12 12 517 736 970 1213 1464 1721 1983
16 16 782 1125 1496 1889 2300 2724 3160
20 20 1056 1530 2051 2610 3200 3815 4451
24 24 1334 1944 2624 3360 4143 4965 5821
28 28 1584 2346 3203 4139 5141 6200 7308
32 32 1830 2754 3803 4957 6200 7520 8906
36 36 2081 3173 4424 5808 7308 8906 10591
40 40 2336 3603 5063 6688 8457 10350 12352
44 44 2595 4040 5717 7593 9643 11844 14179
48 48 2856 4485 6384 8519 10860 13382 16065
7 12 12 540 764 1002 1250 1504 1764 2028
16 16 836 1189 1570 1971 2389 2821 3262
20 20 1141 1632 2169 2742 3345 3971 4617
24 24 1452 2087 2788 3544 4346 5185 6056
28 28 1766 2548 3420 4368 5379 6444 7556
32 32 2050 2988 4044 5198 6438 7751 9128
36 36 2304 3411 4666 6048 7540 9128 10800
40 40 2562 3842 5304 6925 8683 10561 12545
44 44 2822 4280 5958 7826 9861 12044 14356
48 48 3086 4726 6624 8748 11071 13569 16224187 641262 536438
80
129
183
239
299
361
427
496
567
70
111
157
205
255
306
361
417
476
59
94
132
172
213
254
298
343
390
40
62
86
110
135
160
184
210
236
274
309
346
21
33
45
57
70
81
91
101
112
122
30
47
65
82
101
118
113 177 252 337 429 529 635
135
152
169
49
78
108
140
173
205
239
92 142 200 264 334 409 488
103 160 226 300 381 468 560
72 109 150 196 245 297 352
82 125 175 230 289 352 419
53 77 104 133 164 196 230
63 93 127 164 203 245 289
78
31 44 59 75 91 108 125
42 60 81 103 126 151 176
20 29 38 48 58 68
66
77
88
98
109
108
125
142
160
177
150
175
200
226
252 337
245
289
334
381
429
196
230
264
300
529
352
419
488
560
635
297
352
409
468
)
p
A
(fT
cp
2
cp'
cu
cp
2
cp
p
A
cp
2
cp
p
A
Table 3-12 Threshold Torsion
bf
bf
bf
b
h
4hf
(h-h )f
ACI 13.2.4}
hf
≤
≤

For spandrel beam integral with a slab, torsion can be neglected if:
(see Table 3-11 for bf
and hf
)
A simplified sizing equation to neglect torsion effects can be derived based on the limiting factored torsional
moment . Total moment transferred from slab to spandrel = 0.3M0
, (ACI 13.6.3.3) permits
taking the torsional loading from a slab as uniformly distributed along the member. If the uniformly distrib-
uted torsional moments is tu
:
3-28
Acp
2
/pcp Acp
2
/pcp Acp
2
/pcp Acp
2
/pcp Acp
2
/pcp Acp
2
/pcp Acp
2
/pcp
8 12 12 552 781 1022 1272 1529 1791 2057
16 16 879 1241 1630 2039 2463 2900 3347
20 20 1217 1724 2274 2860 3474 4110 4766
24 24 1563 2219 2941 3716 4534 5389 6275
28 28 1912 2723 3622 4594 5628 6715 7847
32 32 2264 3233 4313 5489 6747 8076 9467
36 36 2595 3725 4994 6384 7878 9464 11130
40 40 2856 4160 5636 7263 9020 10893 12868
44 44 3120 4602 6292 8164 10196 12369 14669
48 48 3386 5051 6960 9086 11402 13888 16526
9 12 12 542 774 1018 1272 1532 1797 2066
16 16 910 1280 1676 2091 2521 2962 3413
20 20 1283 1803 2366 2962 3586 4232 4896
24 24 1664 2341 3081 3872 4706 5575 6474
28 28 2050 2888 3812 4807 5863 6970 8120
32 32 2440 3442 4554 5760 7046 8401 9817
36 36 2831 4000 5304 6726 8249 9861 11552
40 40 3224 4561 6061 7701 9467 11343 13317
44 44 3492 5009 6721 8607 10645 12819 15114
48 48 3762 5461 7393 9531 11852 14336 16967
10 12 12 517 749 994 1248 1509 1774 2043
slab
thickness hf
(in.)
Beam
depth h
(in.)
Max.
Slab
width bf
(in.)
Beam width h (in.)
12 16 20 24 28 32 36
Tu
ft kips
Tu
ft kips
Tu
ft kips
Tu
ft kips
Tu
ft kips
Tu
ft kips
Tu
ft kips
507
567
82
135
194
256
321
388
457
526
597
671
71
117
167
220
275
332
390
448
266
304
340
377
61
100
142
186
232
279
326
374
421
468
50
83
117
153
190
228
216
40
66
94
122
151
180
210
240
266
292
31
51
71
93
114
136
158
180
198
134 200 275 359 451 549 653
20 30 39 49 60 70 81
21
36
51
66
81
96
112
127
138
149
113 164 223 287 357 431 509
123 182 249 323 403 489 580
90 128 170 217 267 319 374
103 147 197 252 311 374 440
62 88 116 147 179 213 248
76 108 143 182 222 265 310
35 49 64 81 97 115 132
48 68 90 113 137 162 188
22 31 40 50 60 71 81
16 16 929 1305 1706 2126 2560 3006 3460
20 20 1336 1868 2441 3048 3680 4334 5006
24 24 1754 2449 3205 4012 4860 5742 6653
28 28 2178 3041 3988 5005 6080 7206 8374
32 32 2607 3641 4783 6017 7329 8709 10148
36 36 3038 4245 5587 7043 8599 10242 11962
40 40 3470 4854 6397 8080 9885 11798 13808
44 44 3905 5465 7212 9124 11183 13373 15679
48 48 4219 5962 7926 10086 12421 14912 17545 694
620
167 236 313 399 491 589
137 192 253 319 391 466 546
154 216 285 361 442 529
103 144 189 238 290 344 401
120 168 221 278 340 405 473
69 97 127 159 192 227 263
86 120 158 198 240 285 331
37 52 67 84 101 119 137
53 74 96 120 145 171 198
Table 3-12 Threshold Torsion (Continued)
Tu
< φ ʹfc
bh + bf
hf( )2
2 h + b + bf( )
t u =
0.3Mo
Tu = φ ʹfc
A2
cp
pcp
⎛
⎝
⎜
⎞
⎠
⎟

At the critical section, at distance d away from the support, the critical torsional moment is:
For preliminary sizing of spandrel to neglect torsion, assume d = 0.9h and = 20 per Table 3-1
Therefore
For preliminary sizing of spandrel beam to neglect torsion:
for simplicity:
For one-way slab or one-way joists with spandrel beams, the exterior negative slab moment is equal to wu
˜n
2/24
(ACI 8.3.3). This negative moment can also be expressed as 0.33Mo where Mo = total static span moment for
one-way slab = wu˜2˜n
2/8. Thus for one-way system with spandrel beam, sizing to neglect torsion for the selected
concrete (compressive strength = 4000 psi) reduces to:
3-29
Tu
Tu
tu
˜
+
+
Figure 3-15 Torsional Loading on a Spandrel Beam
Tu
= tu
2
− d
⎛
⎝
⎜
⎞
⎠
⎟ =
0.3Mo
2
− d
⎛
⎝
⎜
⎞
⎠
⎟ = 0.15Mo
1−
2d⎛
⎝
⎜
⎞
⎠
⎟
d
=
0.9h
20h
= 0.05
Tu
= 0.15Mo
(1− 2 × 0.05) = 0.135Mo
h
Tu
≤ φ ʹfc
A2
cp
pcp
0.135Mo
≤ 0.75 4000
(bh + bf
hf
)2
2(h + b + bf
)
(bh + bf
hf
)2
2(h + b + bf
)
>
0.135(Mo
×12,000) × 2
0.75 4000
> 68.2Mo
(bh + bf
hf
)2
2(h + b + bf
)
> 70Mo
(bh + bf
hf
)2
(h + b + bf
)
> 76Mo

The above sizing equations are mixed units: Mo
ft-kips and section dimensions in inches.
Architectural or economic considerations may dictate a smaller spandrel size than that required to neglect
torsion effects. For a specific floor framing system, both architectural and economic aspects of a larger beam
size to neglect torsion versus a smaller beam size with torsion reinforcement (additional closed stirrups at close
spacing combined with longitudinal bars) must be evaluated. If a smaller spandrel with torsion reinforcement is
a more appropriate choice, Section 3.7.2 provides a simple method for the design of the torsion reinforcement.
3.7.1.1 Example: Beam sizing to Neglect Torsion
Determine a spandrel beam size to neglect torsion effects for Building #2, Alternate (1) – slab and column
framing with spandrel beams.
For N-S spandrels: ˜2
= 20 ft
˜1
= 24 ft
˜n
= 24 – (12 + 16)/(2 ϫ 12) = 22.83 ft
wu
= 1.2(136) + 1.6(50) = 243 psf
Mo
= qu
˜2
˜n
2
/8 = 0.243 ϫ 20 ϫ 22.832/8 = 317 ft-kips
For slab thickness hf
= 8.5 in., for monolithic construction (ACI 13.2.4) the portion of slab considered in beam
design is the smaller of:
4hf
h – hf
For preliminary section calculations assume bf
= 3hf
= 3(8.5) = 25.5 in.
Some possible combinations of b and h that will satisfy the requirement for neglecting torsion are shown in Table 3-13.
Clearly large beam sizes are required to neglect torsion effects. It would be more economical to select a smaller
beam and provide torsion reinforcement.
3-30
bh + 25.5( ) 8.5( )( )2
h + b + 25.5( )
= 70(317)

3-31
3.7.2 Beam Design Considering Torsion
It is important for designers to distinguish between two types of torsions: equilibrium torsion and compatibility
torsion. Equilibrium torsion occurs when the torsional resistance is required to maintain static equilibrium.
A simple beam supporting a cantilever along its span is an example of equilibrium torsion. For this case, if
sufficient torsional resistance is not provided, the structure will become unstable and collapse. External loads
have no alternative load path and must be resisted by torsion. Compatibility torsion develops where
redistribution of torsional moments to adjacent members can occur. The term compatibility refers to the
compatibility of deformation between adjacent parts of a structure. As an example, consider a spandrel beam
supporting an exterior slab. As load on the slab increases, so does the negative slab end moment, which induces
torsion in the spandrel beam. The negative slab end moment will be proportional to the torsional stiffness of
the spandrel beam. When the magnitude of the torsional moment exceeds the cracking torque, torsional cracks
spiral around the member, and the cracked torsional stiffness of the spandrel beam is significantly reduced.
As a result, some of the slab negative end moment is redistributed to the slab midspan.
For members in which redistribution of the forces is not possible (equilibrium torsion), the maximum factored
torsional moment, Tu, at the critical section cannot be reduced (Section 11.5.2.1). In this case the design should
be based on Tu
.
For members in a statically indeterminate structure where redistribution of forces can occur (compatibility
torsion), the maximum factored torsional moment at the critical section can be reduced to Tcr
where
. When Tcr/4 < Tu < Tcr, the section should be designed to resist Tu. It is important to
note that the redistribution of internal forces must be considered in the design of the adjoining members.
b (in.)
h (required)
in.
Possible
selection bxh
(in. x in.)
20 68.8
22 60.5
24 54.2
26 49.1
28 45.1 28 x 46
30 41.7 30 x 42
32 38.9 32 x 40
34 36.6 34 x 38
36 34.5 36 x 36
38 32.7 40 x 32
40 31.2
42 29.8
44 28.6
Table 3-13 Beam Dimensions for Neglecting Tension
Tcr = φ4λ ʹfc
A2
cp
pcp
⎛
⎝
⎜
⎞
⎠
⎟

3.7.3 Simplified Design for Torsion Reinforcement
When required, torsion reinforcement must consist of a combination of closed stirrups and longitudinal
reinforcement (ACI 11.5.4). For spandrel beams built integrally with a floor slab system (where reduction of
torsional moment can occur due to redistribution of internal forces after cracking), a maximum torsional
moment of may be assumed for normal weight concrete (ACI 11.5.2.2). The ACI provisions
for members subjected to factored torsional moment Tu and factored shear force Vu for normal weight concrete
can be summarized as follows:
(1) Check if the torsion can be neglected
(ACI 11.5.1)
(2) If torsion cannot be neglected, for members in statically indeterminate structures where redistribution of
forces can occur, calculate the maximum factored torsional moment at the critical section to be considered
(ACI 11.5.2.2):
(ACI 11.5.2.2)
(3) Design the torsional reinforcement for a torque equal to the smaller of the factored torsional moment from
an elastic analysis or the value computed in Step 2. Consider only the spandrel beam reinforcement to resist
torsion. Ignore portion of slab framing in spandrel. If portion of slab is included, torsion reinforcement must
be included in the slab as well.
(4) Check the adequacy of the section dimensions. The cross-sectional dimensions need to be increased if:
(ACI 11.5.3)
where:
Vu
= factored shear force at the section, kips.
ρh
= perimeter of centerline of outermost closed stirrups, in.
Aoh = area enclosed by centerline of the outermost closed stirrups, in2
Vc
= nominal shear strength provided by concrete, kips
bw
= web width, in.
d = distance from extreme compression fiber to centroid of tension reinforcement, in.
(5) Calculate the required area of stirrups for torsion:
(ACI 11.5.3.6)
3-32
φ4λ ʹfc
A2
cp
pcp
⎛
⎝
⎜
⎞
⎠
⎟
Vu
bw
d
⎛
⎝
⎜
⎞
⎠
⎟
2
+
Tu ph
1.7A2
oh
⎛
⎝
⎜
⎞
⎠
⎟
2
≥ φ
Vc
bw
d
+ 8 ʹfc
⎛
⎝
⎜
⎞
⎠
⎟
At
s
=
Tu
2φAo
fy
Tcr ≤ φ λ ʹfc
A2
cp
pcp
⎛
⎝
⎜
⎞
⎠
⎟
Tcr = φ4λ ʹfc
A2
cp
pcp
⎛
⎝
⎜
⎞
⎠
⎟

where:
At
= area of one leg of closed stirrups resisting torsion, in2
Ao
= 0.85Aoh
, in as defined in Step 4
s = spacing of stirrups, in.
(Note that for simplicity the angle of compression diagonals in truss analogy is assumed to be 45°)
(6) Calculate the required area of stirrups for shear:
where:
Av
= Area of shear reinforcement within spacing s (two legs)
(7) Calculate required combined stirrups for shear and torsion (for one leg):
Select the size and spacing of the combined stirrups to satisfy the following conditions:
a.The minimum area of stirrups Av
+2At
(two legs) for the selected concrete with compressive
strength = 4000 psi is 50bw
s/fy
b. The maximum spacing of transverse torsion reinforcement s is the smaller of ph
/8 or 12 in.
(8) Calculate the required additional longitudinal reinforcement A˜ for torsion
(ACI 11.5.3.7)
At/s in the above equation must be taken as the actual amount calculated in Step 5 but not less than 25bw/fy
(ACI 11.5.5.3). The additional longitudinal reinforcement must be distributed around the perimeter of the
closed stirrups with maximum spacing of 12 in.
3-33
Vs
=
Vu
φ
− Vc
Av
s
=
Vs
fy
d
At
s
=
Av
2s
A =
At
s
ph ≥
5 ʹfc
Acp
fy
−
At
s
⎛
⎝
⎜
⎞
⎠
⎟ ph

3-34
3.7.3.1 Example: Design of Torsion Reinforcement
Determine the required combined shear and torsion reinforcement for the E-W spandrel beam of Building #2,
Alternate (1) – slab and column framing with spandrel beams.
For E-W spandrels:
Spandrel size = 12 ϫ 20 in.
d = 20 –2.5 = 17.5 in. = 1.46 ft.
˜n = 24 – (12/12) = 23.0 ft.
Beam weight = 1.2(12 ϫ 20 ϫ 0.150/144) = 0.30 kips/ft
qu from slab = 1.2(136) + 1.6(50) = 243 psf
Tributary load to spandrel (1/2 panel width) = 243 ϫ (20/2) = 2.43 kips/ft
Beam = 0.30 kips/ft
Total wu
= 2.73 kips/ft
Mo
= qu
˜2
˜n
2
/8 = 0.243(24)(18.83)2
/8 = 254.5 ft-kip
where
Tu
= 0.30Mo
/2 = 0.3(254.5) = 38.8 ft-kip
Tu
at distance d = 38.8 – 38.8/24(1.46) = 36.4 ft-kip
Vu
(at the face of support) = 2.73(23)/2 = 31.4 kips
Vu
at distance d = 31.4-2.73(1.46) = 27.4 kips
(1) Check if torsion can be neglected
Acp
= 20(12) + (20-8.5)(8.5) = 337.8 in2
pcp = 20(2) + 12(2) + (20-8.5)(2) = 87 in
= 0.004 = 0.004(337.8)2
/87 = 5.25 ft - kip < 36.4 Torsion must be considered
(2) Calculate the maximum factored torsional moment at the critical section to be considered:
Note: The difference in torsional moments 36.4-20.7 must be redistributed (ACI 11.5.2.2).
This redistribution will result in an increase of the positive midspan moment is slab.
n
= 20 −
12 +16
2(12)
= 18.83ft.
A2
cp
Pcp
⎛
⎝
⎜
⎞
⎠
⎟
Tu = φ4 ʹfc
A2
cp
pcp
⎛
⎝
⎜
⎞
⎠
⎟ = 0.75(4) 4000(337.8)2
/ 87 / 12,000 = 20.7ft − kips < 36.4

(3) Check section adequacy:
Assume the distance to stirrups centroid = 1.75 in. (considering 1.5 in. cover and 0.5 in. stirrups diameter)
ph
= 2(12-2 ϫ 1.75)+2(20-2 ϫ 1.75) = 50 in.
Aoh
= (12-2 ϫ 1.75)(20-2 ϫ 1.75) = 140.3 in2.
Section is adequate
(4) Required stirrups for torsion
Ao
= 0.85Aoh
= 0.8(140.3) = 119.2
(5) Required stirrups for shear
(6) Combined stirrups
maximum stirrups spacing = ph/8 or 12 in. = 50/8 = 6.25 in.
For No. 4 closed stirrups required s = 0.2/0.0325 = 6.15 in.
For No. 3 closed stirrups required s = 0.11/.0325 = 3.4 in.
Use No. 4 closed stirrups @ 6 in.
Area for 2 legs = 0.4 in.2
3-35
Vc
bw
d
= 2 4000 = 126.4 psi
Vu
bw
d
⎛
⎝
⎜
⎞
⎠
⎟
2
+
Tu
ph
1.7A2
oh
⎛
⎝
⎜
⎞
⎠
⎟ =
27.4(1000)
12(17.5)
⎛
⎝
⎜
⎞
⎠
⎟
2
+
20.7(12,000)(50)
1.7(140.3)2
⎛
⎝
⎜
⎞
⎠
⎟
2
= 393.4
2
φ
V
b d
+ 8 ʹfc
⎛
⎝
⎜
⎞
⎠
⎟ = 0.75(126.4 + 8 4000) = 474.3 > 393.4
c
w
At
s
=
Tu
2φAo
fy
=
20.7(12,000)
2(0.75)(119.2)(60,000)
= 0.023in.2
/ in.(one leg)
Vs
=
Vu
φ
− Vc
= 27.4 / 0.75 −126.4(12)(17.5) /1000 = 10 kips
Av
s
=
Vs
fy
d
= 10 / 60 /17.5 = 0.0095 in (two legs)
At
s
+
Av
2s
= 0.023+ 0.0095 / 2 = 0.028in.2
/ in.

3-36
Minimum area of stirrups:
50bw
s/fy
= 50(12)(6)/60,000 = 0.06 in.2
(7) Required additional longitudinal reinforcement:
Place the longitudinal bars around the perimeter of the closed stirrups, spaced not more than 12 in. apart. Locate
one longitudinal bar in each corner of the closed stirrups (ACI 11.5.5.2). For the 20 in. deep beam, one bar is
required at mid-depth on each side face, with 1/3 of the total A˜
required at top, mid-depth, and bottom of the
closed stirrups. A˜
/3 = 1.7/3 = 0.57 in2
. Use 2-No.5 bars at mid-depth (one on each face). Longitudinal bars
required at top and bottom may be combined with the flexural reinforcement.
Details of the shear and torsion reinforcement (at support) are shown in Fig. (3-16).
3.8 EXAMPLES: SIMPLIFIED DESIGN FOR BEAMS AND ONE-WAY SLABS
The following three examples illustrate the use of the simplified design data presented in Chapter 3 for propor-
tioning beams and slabs. Typical floor members for the one-way joist floor system of Building #1 are designed.
A =
At
s
ph = 0.023(50) = 1.15in.2
govern
Minimum =
5 ʹfc
Acp
fy
−
At
s
⎛
⎝
⎜
⎞
⎠
⎟ ph =
5 4000(337.8)
60000
− 0.023( )50 = 0.63in.2
,
20"
81/2"
12"
2-#4
0.38 in.2 + As (required for flexure)
#4 closed stirrups @ 6 in. o.c.
0.38 in.2 + As (required for flexure)
Figure 3-16 Required Shear and Torsion Reinforcement (Example 3.7.2.1)

3-37
3.8.1 Example: Design of Standard Pan Joists for Alternate (1) Floor System
(Building #1)
(1) Data: › = 4000 psi (normal weight concrete, carbonate aggregate)
fy
= 60,000 psi
Floors: LL = 60 psf
DL = 130 psf (assumed total for joists and beams + partitions + ceiling & misc.)
Required fire resistance rating = 1 hour
Floor system – Alternate (1): 30-in. wide standard pan joists
width of spandrel beams = 20 in.
width of interior beams = 36 in.
Note: The special provisions for standard joist construction in ACI 8.13 apply to the pan joist floor
system of Alternate (1).
(2) Determine the factored shears and moments using the approximate coefficients of ACI 8.3.3.
Factored shears and moments for the joists of Alternate (1) are determined in Chapter 2, Section 2.3.2
(Figure 2-8).
wu
= [1.2(130) + 1.6(60) = 252 psf] ϫ 3 ft = 756 plf
Note: All shear and negative moment values are at face of supporting beams.
Vu
@ spandrel beams = 10.5 kips
Vu
@ first interior beams = 12.0 kips
-Mu
@ spandrel beams = 24.0 ft-kips
+Mu
@ end spans = 41.1 ft-kips
-Mu
@ first interior beams = 56.3 ft-kips
+Mu
@ interior spans = 34.6 ft-kips
-Mu
@ interior beams = 50.4 ft-kips
(3) Preliminary size of joist rib and slab thickness
From Table 3-1: depth of joist h = ñ
/18.5 = (27.5 ϫ 12)/18.5 = 17.8 in.
where ñ
(end span) = 30 – 1.0 – 1.5 = 27.5 ft (governs)
ñ
(interior span) = 30 – 3 = 27.0 ft
From Table 10-1, required slab thickness = 3.2 in. for 1-hour fire resistance rating. Also, from
ACI 8.13.6.1:
Minimum slab thickness > 30/12 = 2.5 in. > 2.0 in.

3-38
Try 16 in. pan forms * + 31/2
-in. slab
h = 19.5 in. > 17.8 in. O.K.
slab thickness = 3.5 in. > 3.2 in. O.K.
(4) Determine width of joist rib
(a) Code minimum (ACI 8.13.2):
bw
> 16/3.5 = 4.6 in. > 4.0 in.
(b) For flexural strength:
where d = 19.5 – 1.25 = 18.25 in. = 1.52 ft
Check for fire resistance: from Table 10-4 for restrained members, the required cover for a fire
resistance rating of 1 hr = 3
/4
-in. for joists.
(c) For shear strength
Vu
@ distance d from support face = 12.0 – 0.76(1.52) = 10.84 kips
φVc
= 1.1(0.095bw
d) = 0.11bw
d**
= 10.84/(0.11 x 18.25) = 5.40 in.
Use 6 in. wide joists (see Table 9-3 and Fig. 3-17).
(5) Determine Flexural Reinforcement
(a) Top bars at spandrel beams:
Distribute bars uniformly in top slab:
As
= 0.33/3 = 0.11 in.2
/ft
bw =
20Mu
d2
=
20(57.5)
18.252
= 3.45 in.
* See Table 9-3 for standard form dimensions for one-way joists.
** For standard joist ribs conforming to ACI 8.13 a 10% greater shear strength fVc is allowed. Also, minimum shear
reinforcement is not required (see ACI 11.4.6).
∴bw
As
=
Mu
4d
=
24.0
4(18.25)
= 0.33in.2

Maximum bar spacing for an interior exposure and 3
/4
-in. cover will be controlled by
provisions of ACI 7.6.5
smax
= 3h = 3(3.5) = 10.5 in. < 18 in.
From Table 3-6: Use No.3 @ 10 in. (As
= 0.13 in.2
/ft)
(b) Bottom bars in end spans:
As
= 41.1/(4 ϫ 18.25) = 0.56 in.2
Check rectangular section behavior:
a = Asfy/0.85 ›be = (0.56 ϫ 60)/(0.85 ϫ 4 ϫ 36) = 0.27 in. < 3.5 in. O.K.
From Table 3-5: Use 2-No.5 (As
= 0.62 in.2)
Check ρ = As
/bw
d = 0.56/(6 ϫ 18.25) = 0.005 > ρmin
= 0.0033 O.K.
(c) Top bars at first interior beams:
As
= [56.3/ (4 ϫ 18.25) = 0.77 in.2]/3 = 0.26 in.2/ft
From Table 3-6, with smax
= 10.5 in.:
Use No.5 @ 10 in. (As = 0.37 in.2/ft)
3-39
12
Joist top bars Slab bars
Bottom bars
30" clear30" clear 6"
16"3.5"
36"
Figure 3-17 Joist Section (Example 3.8.1)
c

3-40
(d) Bottom bars in interior spans:
As
= 34.6/(4 ϫ 18.25) = 0.47 in.2
From Table 3-5: Use 2-No.5 (As
= 0.62 in.2)
(e) Top bars at interior beams:
As
= [50.4/(4 ϫ 18.25) = 0.69 in.2]/3 = 0.23 in.2/ft
From Table 3-6, with smax
= 10.5 in.:
Use No.4 @ 10 in. (As
= 0.24 in.2
/ft)
(f) Slab reinforcement normal to ribs (ACI 8.13.6.2):
Slab reinforcement is often located at mid-depth of slab to resist both positive and negative moments.
Use Mu
= wu
ñ
2
/12 (see Fig. 2-5)
= 0.19(2.5)2
/12 = 0.10 ft-kips
where wu
= 1.2(44 + 30) + 1.6(60) = 185 psf = 0.19 kips/ft
ñ
= 30 in. = 2.5 ft (ignore rib taper)
With bars on slab centerline, d = 3.5/2 = 1.75 in.
As
= 0.10/4(1.75) = 0.015 in.2
/ft (but not less than required temperature and shrinkage
reinforcement)
From Table 3-4, for a 31
/4-
in. slab: Use No.3 @ 16 in.
Note: For slab reinforcement normal to ribs, space bars per ACI 7.12.2.2 at 5h or 18 in.
Check for fire resistance: From Table 10-3, required cover for fire resistance rating of
1 hour = 3
/4-in. O.K.
Figure 3-18 Reinforcement Details for 30 in. Standard Joist Floor System [Building #1—Alternate (1)]
#3 @ 16" (typ)
1-#5
1-#5
7'-11"
#3 @ 10"
6" 0"
1'-8"
8"
9'-9" 9'-9"
#4 @ 10"
1-#5
1-#53'-5" 3'-5"
3'-0"
16"
3.5"
30'-0"
ñ = 27.0'
Sym about CL
End Span Interior Span (Typ)
ñ = 27.5'
30'-0"

3-41
(6) Reinforcement details shown in Fig. 3-18 are determined directly from Fig. 8-5*. Note that
one of the No.5 bars at the bottom must be continuous or be spliced over the support with a
Class B tension splice, and be terminated with a standard hook at the non-continuous supports
for structural integrity (ACI 7.13.2.1).
3.8.2 Example: Design of Wide-Module Joists for Alternate (2) Floor System
(Building #1)
fy = 60,000 psi
Floors: LL = 60 psf
Fire resistance rating = assume 2 hours
Floor system – Alternate (2): Assume joists on 6 ft – 3 in. centers (53” standard forms** )
width of spandrel beams = 20 in.
width of interior beams = 36 in.
Note: The provisions for standard joist construction in ACI 8.13 do not apply for the wide-module joist
system (clear spacing between joists > 30 in.). Wide-module joists are designed as beams and one-way
slabs (ACI 8.13.4).
(2) Determine factored shears and moments using the approximate coefficients of ACI 8.3.3
(see Figs. 2-3, 2-4, and 2-7).
wu = 1.2(130) + 1.6(60)*** = 252 psf ϫ 6.25 = 1575 plf, say 1600 plf
Vu @ spandrel beams = 1.6(27.5)/2 = 22.0 kips
Vu @ first interior beams = 1.15(22.0) = 25.3 kips
Vu @ interior beams = 1.6(27.0)/2 = 21.6 kips
-Mu @ spandrel beams = 1.6(27.5)2/24 = 50.4 ft-kips
+Mu @ end spans = 1.6(27.5)2/14 = 86.4 ft-kips
-Mu @ first interior beams = 1.6(27.25)2/10 = 118.8 ft-kips
+Mu @ interior spans = 1.6(27)2/16 = 72.9 ft-kips
-Mu @ interior beams = 1.6(27)2/11 = 106 ft-kips
* The bar cut-off points shown in Fig. 8-5 are recommended for one-way joist construction. The reader may consider
determining actual bar lengths using the provisions in ACI 12.10.
** See Table 9-3 for standard form dimensions.
***No live load reduction permitted: since the influence area = 12.5 ϫ 30 = 375 sq ft < 400 sq ft.

3-42
(3) Preliminary size of joists (beams) and slab thickness
From Table 3-1: depth of joist h = ñ
/18.5 = (27.5 ϫ 12)/18.5 = 17.8 in.
where ñ
ñ
(interior span) = 30 – 3 = 27.0 ft
Check for fire resistance: from Table 10-1, required slab thickness = 4.6 in. for 2-hour rating.
Try 16 in. pan forms + 41/2
-in. slab
h = 20.5 in. > 17.8 in. O.K.
(4) Determine width of joist rib
(a) For moment strength:
where d = 20.5 – 2.5 = 18.0 in. = 1.50 ft
Check for fire resistance: for joists designed as beams, specified cover per ACI 7.7 = 1.5 in.
From Table 10-4, the required cover for restrained beams for a fire resistance rating of
2 hr = 1
/4
-in. < 1.5 in. O.K.
(b) For shear strength
Vu
@ distance d from support face = 25.3 – 1.6(1.50) = 22.9 kips
φVc
/2 = 0.048bw
d*
bw
= 22.9/(0.048 ϫ 18.0) = 26.5 in.
Use 9 in.-wide joists (standard width) and provide stirrups where required. A typical
cross-section of the joist is shown in Fig. 3-19.
(5) Determine Flexural Reinforcement
(a) Top bars at spandrel beams:
Distribute bars uniformly in top slab according to ACI 10.6.6.
* For joists designed as beams, the 10% increase in φVc
is not permitted. Also, minimum shear reinforcement is required when
Vu > φVc/2.
bw =
20Mu
d2
=
20(118.8)
18.02
= 7.33in.
As
=
Mu
4d
=
50.4
4(18.0)
= 0.70 in.2

3-43
Effective flange width: (30 ϫ 12)/10 = 36 in. (governs
6.25 ϫ 12 = 75 in.
9 + 2(8 ϫ 4.5) = 81 in.
As
= 0.70/3 = 0.23 in.2/ft
maximum bar spacing for No.4 bars,
smax
= 3h = 3(4.5) = 13.5 in. < 18 in.
From Table 3-6: Use No. 4 @ 9 in. (As
= 0.27 in.2
/ft)
As
= 86.4/(4 ϫ 18.0) = 1.2 in.2
a = As
fy
/0.85 › be
= (1.2 ϫ 60)/(0.85 ϫ 4 ϫ 36) = 0.59 in. < 4.5 in. O.K.
From Table 3-5: Use 2-No. 7 (As
= 1.20 in.2
)
Check ρ = As/bwd = 1.2/(9 ϫ 18.0) = 0.0074 > ρmin = 0.0033 O.K.
The 2-No.7 bars satisfy all requirements for minimum and maximum number of bars
in a single layer.
(c) Top bars at first interior beams:
As
= [118.8/(4 ϫ 18.0) = 1.65 in.2]/3 = 0.55 in.2/ft
From Table 3-6, assuming No.7 bars (smax
= 10.8 in. from Table 3-6),
Use No. 7 @ 10 in. (As = 0.72 in.2
/ft)
Stirrups
Bottom bars
9"
12
16"4.5"
Slab barsJoist top bar
75"
11/2"
11/2"
Figure 3-19 Joist Section

(d) Bottom bars in interior spans:
As
= 72.9/(4 ϫ 18.0) = 1.01 in.2
From Table 3-5: Use 2 No. 7 (As
= 1.20 in.2)
(e) Top bars at interior beams:
As = [106/(4 ϫ 18) = 1.47 in.2
]/3 ft = 0.50 in.2
/ft
From Table 3-6, with smax = 11.6 in.
Use No. 6 @ 10 in. (As = 0.59 in.2/ft)
(f) Slab reinforcement normal to joists:
Use Mu = wu˜n
2
/12 (see Fig. 2-5)
= 0.20(6.25)2
/12 = 0.65 ft-kips
where wu = 1.2(56 + 30) + 1.6(60) = 200 psf = 0.20 kips/ft
Place bars on slab centerline: d = 4.5/2 = 2.25 in.
As
= 0.65/4(2.25) = 0.07 in.2
/ft
(but not less than required temperature reinforcement).
From Table 3-4, for a 41
⁄2 in. slab: Use No. 3 @ 13 in. (As
= 0.10 in.2
/ft)
Check for fire resistance: from Table 10-3, for restrained members, required cover for
fire resistance rating of 2 hours = 3
⁄4 in. O.K.
(6) Reinforcement details shown in Fig. 3-20 are determined directly from Fig. 8-3(a).*
For structural integrity (ACI 7.13.2.3), one of the No. 7 and No. 8 bars at the bottom must be
spliced over the support with a Class B tension splice; the No. 8 bar must be terminated with
a standard hook at the non-continuous supports.
(7) Design of Shear Reinforcement
(a) End spans:
Vu at face of interior beam = 25.3 kips
Vu at distance d from support face = 25.3 – 1.6(1.50) = 22.9 kips
3-44
* The bar cut-off points shown in Fig. 8-3(a) are recommended for beams without closed stirrups. The reader may consider deter-
mining actual bar lengths using the provisions in ACI 12.10.

Use average web width for shear strength calculations
bw
= 9 + (20.5/12) = 10.7 in.
(φVc
+ φVs
)max
= 0.48 bw
d = 0.48(10.7)(18.0) = 92.4 kips
φVc
= 0.095 bw
d = 0.095(10.7)(18.0) = 18.3 kips
φVc/2 = 0.048 bwd = 9.3 kips
Beam size is adequate for shear strength since 92.3 kips > 22.9 kips.
Since Vu
> φVc
, more than minimum shear reinforcement is required. Due to the sloping face of the
joist rib and the narrow widths commonly used, shear reinforcement is generally a one-legged stirrup
rather than the usual two. The type commonly used is shown in Fig. 3-21. The stirrups are attached to
the joist bottom bars.
Single leg No. 3 @ d/2 is adequate for full length where stirrups are required. Length over which
stirrups are required: (25.3 – 9.3)/1.8 = 8.9 ft. Stirrup spacing s = d/2 = 18.0/2 = 9.0 in.
Check Av(min)
= 50 bw
s/fy
= 50 x 10.7 ϫ 9.0/60,000 = 0.08 in.2
Single leg No.3 stirrup O.K.
Use 14-No. 3 single leg stirrups @ 9 in. Use the same stirrup detail at each end of all joists.
3-45
Interior Span (typ)
Class B splice
End Span
#3 @ 9" single leg
stirrups
#3 @ 9" single leg
stirrups#3 @ 9" single leg stirrups
Single leg
#3 stirrups
#7 @ 10" (1st interior)
#6 @ 10" (interior)
10'-8" 10'-8"
1-#7
1-#8
1-#7
1-#7
#3 @ 13"
16"
4.5"
10'-6"
10'-6" 10'-6"
30'-0" 30'-0"
3'-5" 3'-5"3'-0"
0"
6"
8"
1'-8"
7'-11"
4 @ 9"
Sym about CL
Figure 3-20 Reinforcement Details for 6 ft-3 in. Wide-Module Joist Floor System [Building #1—Alternate (2)]
Stirrups
Alternate hook direction at
every other stirrup
Figure 3-21 Stirrup Detail

3.8.3 Example: Design of the Support Beams for the Standard Pan Joist Floor along
a Typical N-S Interior Column Line (Building #1)
fy
= 60,000 psi
Floors: LL = 60 psf
Required fire resistance rating = 1 hour (2 hours for Alternate (2)).
Preliminary member sizes
width of interior beams = 36 in. 2 ϫ depth – 2 ϫ 19.5 = 39.0 in.
The most economical solution for a pan joist floor is making the depth of the supporting beams equal to
the depth of the joists. In other words, the soffits of the beams and joists should be on a common plane.
This reduces formwork costs sufficiently to override the savings in materials that may be accomplished
by using a deeper beam. See Chapter 9 for a discussion on design considerations for economical
formwork. The beams are often made about twice as wide as they are deep. Overall joist floor depth =
16 in. + 3.5 in. = 19.5 in. Check deflection control for the 19.5 in. beam depth. From Table 3-1:
h = 19.5 in. > /18.5 = (28.58 ϫ 12)/18.5 = 18.5 in. O.K.
where ˜n
˜n
(interior span) = 30 – 1.50 = 28.50 ft.
(2) Determine factored shears and moments from the gravity loads using the approximate coefficients
(see Figs. 2-3, 2-4, and 2-7).
Check live load reduction. For interior beams:
KLLAT = 2(30 ϫ 30) = 1800 sq ft > 400 sq ft
L = 60(0.25 + 15/ ) = 60(0.604)* = 36.2 psf > 50% Lo
DL = 130 ϫ 30 ϫ = 3.9 klf
LL = 36.2 ϫ 30 ϫ = 1.09 klf
wu
= [1.20(130) + 1.6(36.2) = 214 psf] ϫ 30 ft = 6.4 klf
1
1000
1
1000
1800
* For members supporting one floor only, maximum reduction = 0.5 (see Table 2-5).
3-46

Vu
@ exterior columns = 6.4(28.58)/2 = 91.5 kips
Vu
@ first interior columns = 1.15(91.5) = 105.3 kips
Vu
@ interior columns = 6.4(28.5)/2 = 91.2 kips
-Mu @ exterior columns = 6.4(28.58)2
/16 = 326.7 ft-kips
+Mu @ end spans = 6.4(28.58)2/14 = 373.4 ft-kips
-Mu @ first interior columns = 6.4(28.58)2/10 = 522.8 ft-kips
+Mu @ interior span = 6.4(28.50)2/16 = 324.9 ft-kips
(3) Design of the column line beams also includes consideration of moments and shears due to wind.
The wind load analysis for Building #1 is summarized in Fig. 2-13.
Note: The reduced load factor (0.5) permitted for load combinations including the wind effect
(ACI 9-3 and 9-4) is in most cases, sufficient to accommodate the wind forces and moments without
an increase in the required beam size or reinforcement (i.e., the load combination for gravity load only
will usually govern for proportioning the beam).
(4) Check beam size for moment strength
Preliminary beam size = 19.5 in. ϫ 36 in.
For negative moment section:
where d = 19.5 – 2.5 = 17.0 in. = 1.42 ft
For positive moment section:
bw
= 20 (373.4)/172
= 25.8 in. < 36 in.
Check minimum size permitted with ρ = 0.0206:
bw = 14.6(522.8)/172
= 26.4 in. < 36 in. O.K.
Use 36 in. wide beam and provide slightly higher percentage of reinforcement (ρ > 0.5 ρmax
) at
interior columns.
Check for fire resistance: from Table 10-4, required cover for fire resistance rating of 4 hours or less
= 3
⁄4 in. < provided cover. O.K.
bw =
20Mu
d2
=
20(522.8)
172
= 36.2 in. > 36 in.
3-47

(5) Determine flexural reinforcement for the beams at the 1st floor level
(a) Top bars at exterior columns
Check governing load combination:
• gravity loads
Mu = 326.7 ft-kips ACI Eq. (9-2)
• gravity + wind load
Mu = 1.2(3.9)(28.58)2/16 + 0.8(99.56) = 317.9 ft-kips ACI Eq. (9-3)
or
Mu = 1.2(3.9)(28.58)2
/16 + 0.5(1.09)(28.58)2
/16
+ 1.6(99.56) = 426.1 ft-kips ACI Eq. (9-4)
• also check for possible moment reversal due to wind moment:
Mu = 0.9(3.9)(28.58)2
/16 ± 1.6(99.56) = 338.5 klf, 19.9 ft-kips ACI Eq. (9-6)
From Table 3-5: Use 8-No. 8 bars (As
= 6.32 in.2
)
minimum n = 36[1.5 + 0.5 + (1.0/2)]2
/57.4 = 4 bars < 8 O.K.
Check ρ = As
/bd = 6.32/(36 ϫ 17) = 0.0103 > ρmin
= 0.0033 O.K.
As
= 373.4/4(17) = 5.49 in.2
Use 8-No. 8 bars (As
= 6.32 in.2)
(c) Top bars at interior columns:
• gravity load only
Mu
= 522.8 ft-kips ACI Eq. (9-2)
• gravity + wind loads:
Mu = 1.2(3.9)(28.54)2/10 + 0.8(99.56) = 460.8 ft-kips ACI Eq. (9-3)
or
Mu
= 1.2(3.9)(28.54)2/10 + 0.5(1.09)(28.54)2/10
+ 1.6(99.56) = 584.9 ft-kips ACI Eq. (9-4)
As
=
Mu
4d
=
426.1
4(17)
= 6.27 in.2
3-48

3-49
As
= 584.9/4(17) = 8.6 in.2
= 8.6 in.2)
(d) Bottom bars in interior span:
As
= 324.9/4(17) = 4.78 in.2
Use 7-No. 8 bars (As = 5.53 in.2
)
(6) Reinforcement details shown in Fig. 3-22 are determined directly from Fig. 8-3(a).* Provided 2-No. 5
top bars within the center portion of all spans to account for any variations in required bar lengths due
to wind effects.
Since the column line beams are part of the primary wind-force resisting system. ACI 12.11.2 requires
at least one-fourth the positive moment reinforcement to be extended into the supporting columns and
be anchored to develop full fy at face of support. For the end spans: As
/4 = 8/4 = 2 bars. Extend 2-No. 8
center bars anchorage distance into the supports:
• At the exterior columns, provide a 90° standard end-hook (general use). From Table 8-5, for
No. 8 bar:
˜n = 14 in. = 16 – 2 = 14 in. O.K.
• At the interior columns, provide a Class B tension splice (ACI 13.2.4). Clear space between
No. 8 bars = 3.4 in. = 3.4db. From Table 8-2, length of splice = 1.0 ϫ 30 = 30 in. (ACI 12.15).
(7) Design of Shear Reinforcement
Design shear reinforcement for the end span at the interior column and use the same stirrup
requirements for all three spans.
• gravity load only
Vu = at interior column = 105.5 kips (governs)
• gravity + wind loads:
Vu
= 1.2(1.15)(3.9)(28.54)/2 + 0.8(6.64) = 82.1 kips
or
Vu = 1.2(1.15)(3.9)(28.54)/2 + 0.5(1.15)(1.09)(28.54)/2 + 1.6(6.64) = 95.7 kips
* The bar cut-off points shown in Fig. 8-3(a) are recommended for beams without closed stirrups. The reader may consider
determining actual bar lengths using the provisions in ACI 12.10.

• wind only at span center
Vu
= 1.6(6.64) = 10.6 kips
Vu
@ face of column = 105.5 kips
Vu
at distance d from column face = 105.5 – 6.4(1.42) = 96.4 kips
(φVc + φVs)max = 0.48 bwd = 0.48(36)17 = 293.8 kips > 96.4 kips O.K.
φVc = 0.095 bwd = 0.095(36)17 = 58.1 kips
φVc/2 = 29.1 kips
3-50
7'-10"
2"
clear
6"
10'-4" 10'-4"
30'-0" 30'-0"
2-#82-#8
2-#5 2-#511-#8
5-#85-#8
8-#8
Class B
tension splice
18"
1'-3" 1'-3"
4'-3" 4'-3"
16"
0"
1.5" clear
to stirrups
11-#8 at interior columns
8-#8 at eterior columns
36"
16"3.5"12
#4 U-stirrups 8-#8 in end spans
7-#8 in interior spans
Figure 3-22 Reinforcement Details for Support Beams along N-S Interior Column Line

Length over which stirrups are required: (105.5 – 29.1)/6.4 = 11.9 ft
φVs
(required) = 96.4 – 58.1 = 38.3 kips
Try No. 4 U-stirrups
From Fig. 3-4, use No.4 @ 8 in. over the entire length where stirrups are required (see Fig. 3-23).
3-51
#4 U-stirrups
Face of
column
18 @ 8" = 12'-0"2"
Figure 3-23 Stirrup Spacing Layout

References
3.1 Fling. R.S., “Using ACI 318 the Easy Way,” Concrete International, Vol. 1, No. 1, January 1979.
3.2 Pickett, C., Notes on ACI 318-77, Appendix A – Notes on Simplified Design, 3rd Edition, Portland Cement
Association, Skokie, Illinois
3.3 Rogers, P., “Simplified Method of Stirrup Spacing,” Concrete International, Vol. 1, No. 1, January 1979.
3.4 Fanella D.A, “Time-Saving DesignAids for Reinforced Concrete – Part 1”, Structural Engineer,August 2001
3.5 Fanella D. A, “Time-Saving Design Aids for Reinforced Concrete - Part 2 Two-Way Slabs”, Structural
Engineer, October 2001
3.6 Fanella D. A, “Time-Saving Design Aids for Reinforced Concrete - Part 3 Columns & Walls”, Structural
Engineer, November 2001
3.7 PSI – Product Services and Information, Concrete Reinforcing Steel Institute, Schaumburg, Illinois.
(a) Bulletin 7901A Selection of Stirrups in Flexural Members for Economy
(b) Bulletin 7701A Reinforcing Bars Required – Minimum vs. Maximum
(c) Bulletin 7702A Serviceability Requirements with Grade 60 Bars
3.8 Notes on ACI 318-08, Chapter 6, “General Principles of Strength Design,” EB708, Portland Cement
Association, Skokie, Illinois, 2008.
3.9 Design Handbook in Accordance with the Strength Design Method of ACI 318-89: Vol. 1–Beams, Slabs,
Brackets, Footings, and Pile Caps, SP-17(91), American Concrete Institute, Detroit, 1991.
3.10 CRSI Handbook, Concrete Reinforcing Steel Institute, Schaumburg, Illinois, 10th Edition, 2008.
3.11 ACI Detailing Manual – American Concrete Institute, Detroit, 1994.
3-52

3-53

4-1
Chapter 4
Simplified Design for Two-Way Slabs
4.1 INTRODUCTION
Figure 4-1 shows the various types of two-way reinforced concrete slab systems in use at the present time.
A solid slab supported on beams on all four sides [Fig. 4-1(a)] was the original slab system in reinforced
concrete. With this system, if the ratio of the long to the short side of a slab panel is two or more, load transfer
is predominantly by bending in the short direction and the panel essentially acts as a one-way slab. As the ratio
of the sides of a slab panel approaches unity (square panel), significant load is transferred by bending in both
orthogonal directions, and the panel should be treated as a two-way rather than a one-way slab.
As time progressed and technology evolved, the column-line beams gradually began to disappear. The resulting
slab system, consisting of solid slabs supported directly on columns, is called a flat plate [Fig. 4-1(b)]. The flat
plate is very efficient and economical and is currently the most widely used slab system for multistory
residential and institutional construction, such as motels, hotels, dormitories, apartment buildings, and
hospitals. In comparison to other concrete floor/roof systems, flat plates can be constructed in less time and
with minimum labor costs because the system utilizes the simplest possible formwork and reinforcing steel
layout. The use of flat plate construction also has other significant economic advantages. For instance, because
of the shallow thickness of the floor system, story heights are automatically reduced resulting in smaller overall
heights of exterior walls and utility shafts, shorter floor to ceiling partitions, reductions in plumbing, sprinkler
and duct risers, and a multitude of other items of construction. In cities like Washington, D.C., where the
maximum height of buildings is restricted, the thin flat plate permits the construction of the maximum number
of stories on a given plan area. Flat plates also provide for the most flexibility in the layout of columns,
partitions, small openings, etc. Where job conditions allow direct application of the ceiling finish to the flat
plate soffit, (thus eliminating the need for suspended ceilings), additional cost and construction time savings
are possible as compared to other structural systems.
The principal limitation on the use of flat plate construction is imposed by punching shear around the columns
(Section 4.4). For heavy loads or long spans, the flat plate is often thickened locally around the columns
creating either drop panels or shear caps. When a flat plate is equipped with drop panels or shear caps, it is
called a flat slab [Fig. 4-1(c)]. Also, for reasons of shear capacity around the columns, the column tops are
sometimes flared, creating column capitals. For purposes of design, a column capital is part of the column,
whereas a drop panel is part of the slab.
Waffle slab construction [Fig. 4-1(d)] consists of rows of concrete joists at right angles to each other with solid
heads at the columns (to increase punching shear resistance). The joists are commonly formed by using
standard square dome forms. The domes are omitted around the columns to form the solid heads acting as drop
panels. Waffle slab construction allows a considerable reduction in dead load as compared to conventional flat

4-2
(a) Two-Way Beam Supported Slab
(b) Flat Plate
(c) Flat Slab
(d) Waffle Slab (Two-Way Joist Slab)
Figure 4-1 Types of Two-Way Slab Systems

4-3
Chapter 4 • Simplified Design for Two-Way Slabs
slab construction. Thus, it is particularly advantageous where the use of long span and/or heavy loads is desired
without the use of deepened drop panels or support beams. The geometric shape formed by the joist ribs is often
architecturally desirable.
Discussion in this chapter is limited largely to flat plates and flat slabs subjected only to gravity loads.
4.2 DEFLECTION CONTROL—MINIMUM SLAB THICKNESS
Minimum thickness/span ratios enable the designer to avoid extremely complex deflection calculations in
routine designs. Deflections of two-way slab systems need not be computed if the overall slab thickness meets
the minimum requirements specified in ACI 9.5.3. Minimum slab thicknesses for flat plates, flat slabs (and
waffle slabs), and two-way slabs, based on the provisions in ACI 9.5.3, are summarized in Table 4-1, where ˜n
is the clear span length in the long direction of a two-way slab panel. The tabulated values are the controlling
minimum thicknesses governed by interior or exterior panels assuming a constant slab thickness for all panels
making up a slab system.4.1
Practical spandrel beam sizes will usually provide beam-to-slab stiffness ratios, αf
,
greater than the minimum specified value of 0.8; if this is not the case, the spandrel beams must be ignored in
computing minimum slab thickness. A standard size drop panel that would allow a 10% reduction in the
minimum required thickness of a flat slab floor system is illustrated in Fig. 4-2. Note that a larger size and depth
drop may be used if required for shear strength; however, a corresponding lesser slab thickness is not permitted
unless deflections are computed.
Table 4-1 and Figure 4-3 gives the minimum slab thickness h based on the requirements given in ACI 9.5.3;
αfm
is the average value of αf
(ratio of flexural stiffness of beam to flexural stiffness of slab) for all beams on
the edges of a panel, and β is the ratio of clear spans in long to short direction.
For design convenience, minimum thicknesses for the six types of two-way slab systems listed in Table 4-1 are
plotted in Fig. 4-4.
h
1.25h
˜/6˜/6
Figure 4-2 Drop Panel Details (ACI 13.3.7)

4-4
Table 4-1 Minimum Thickness for Two-Way Slab Systems
Two-Way Slab System Minimum h
Flat Plate, exterior panel — < 2 n /30
Flat Plate, interior panel and
exterior panel with edge beam1
[Min. h = 5 in.] — < 2 n /33
Flat Slab2 — < 2 n /33
[Min. h = 4 in.] — < 2 n /36
< 0.2
1.0
< 2
1
n /30
n /33
Two-Way Beam-Supported Slab3
> 2.0
2
1
2
n /36
n /37
n /44
< 0.2
1.0
< 2
1
n /33
n /36
Two-Way Beam-Supported Slab1,3
> 2.0
2
1
2
n /40
n /41
n /49
1 Spandrel beam-to-slab stiffness ratio > 0.8 (ACI 9.5.3.3)
2 Drop panel length > /3, depth > 1.25h (ACI 13.3.7)
3 Min. h = 5 in. for < 2.0; min. h = 3.5 in. for > 2.0 (ACI 9.5.3.3)
βαfm
αm αm
αf
˜
˜
˜
˜
˜
˜
˜
˜
˜
˜
˜
˜
˜
˜
˜
Flat Slab, interior panel and
exterior panel with edge beam1
n /33˜ n /33˜
n /33˜ n /30˜
n /30˜ n /30˜
n /36˜ n /36˜
n /36˜ n /33˜
n /33˜ n /33˜
Edge
Beams
Edge
Beams
αf for edge beam ≥ 0.8
˜n = the longer clear span
Flat Plates Flat Slabs
Figure 4-3 Minimum Thickness for Flat Plates and Flat Slab (Grade 60 Reinforcing Steel)

4-5
3
4
5
6
7
8
9
10
11
12
13
14
10 15 20 25 30 35 40
Longer Clear Span (ft)
*Spandrel beam-to-slab stiffness ratio αf ≥ 0.8
**αfm > 2.0
Two-way beam-supported
slab (β = 1) ***
Two-way beam-supported
slab (β = 2) ***
Flat slab with
spandrel beams*
Flat plate with spandrel
beams* and flat slab
Flat plate
Figure 4-4 Minimum Slab Thickness for Two-Way Slab Systems

4.3 TWO-WAY SLAB ANALYSIS BY COEFFICIENTS—DIRECT DESIGN METHOD
For gravity loads, ACI Chapter 13 provides two analysis methods for two-way slab systems: the Direct Design
Method (ACI 13.6) and the Equivalent Frame Method (ACI 13.7). The Equivalent Frame Method, using
member stiffnesses and complex analytical procedures, is not suitable for hand calculations. Only the Direct
Design Method, using moment coefficients, will be presented in this Chapter.
The Direct Design Method applies when all of the conditions illustrated in Fig. 4-5 are satisfied (ACI 13.6.1):
• There must be three or more continuous spans in each direction.
• Slab panels must be rectangular with a ratio of longer to shorter span (c/c of supports) not greater than 2.
• Successive span lengths (c/c of supports) in each direction must not differ by more than one-third of the
longer span.
• Columns must not be offset more than 10% of the span (in direction of offset) from either axis between
centerlines of successive columns.
• Loads must be due to gravity only and must be uniformly distributed over the entire panel.
• The unfactored live load must not be more than 2 times the unfactored dead load (L/D ≤ 2).
• For two-way slabs, relative stiffnesses of beams in two perpendicular directions must satisfy the minimum
and maximum requirements given in ACI 13.6.1.6.
• Redistribution of moments by ACI 8.4 must not be permitted.
4-6
˜1
Uniformly Distributed Loading (L/D ≤ 2)
˜1 ˜1(≥ 2/3)
Three or More Spans
˜2
Column offset
Rectangular slab
panels (2 or less: 1)
≤ ˜2/10
Figure 4-5 Conditions for Analysis by the Direct Design Method

In essence, the Direct Design Method is a three-step analysis procedure. The first step is the calculation of the
total design moment Mo
for a given panel. The second step involves the distribution of the total moment to the
negative and positive moment sections. The third step involves the assignment of the negative and positive
moments to the column strips and middle strips.
For uniform loading, the total design moment Mo
for a panel is calculated by the simple static moment
expression, ACI Eq. (13-4):
Mo
= qu
˜2
ñ
2/8
where qu
is the factored combination of dead and live loads (psf), qu
= 1.2 wd
+ 1.6 w˜
. The clear span ñ
is defined in a straightforward manner for slabs supported on columns or other supporting elements of
rectangular cross section (ACI 13.6.2.5). Circular or regular polygon shaped supports must be treated as square
supports with the same area (see ACI Fig. R13.6.2.5). The clear span starts at the face of support and must not
be taken less than 65% of the span center-to-center of supports (ACI 13.6.2.5). The span ˜2
is simply the span
transverse to ñ
; however, when the panel adjacent and parallel to an edge is being considered, the distance from
edge of slab to panel centerline is used for ˜2
in calculation of Mo
(ACI 13.6.2.4).
Division of the total panel moment Mo
into negative and positive moments, and then into column and
middle strip moments, involves direct application of moment coefficients to the total moment Mo
. The moment
coefficients are a function of span (interior or exterior) and slab support conditions (type of two-way slab
system). For design convenience, moment coefficients for typical two-way slab systems are given in Tables
4-2 through 4-6. Tables 4-2 through 4-5 apply to flat plates or flat slabs with various end support conditions.
Table 4-6 applies to two-way slabs supported on beams on all four sides. Final moments for the column strip
and middle strip are computed directly using the tabulated values. All coefficients were determined using the
appropriate distribution factors in ACI 13.6.3 through 13.6.6.
4-7
End Span Interior Span
Slab Moments
1
Exterior
Negative
2
Positive
3
First Interior
Negative
4
Positive
5
Interior
Negative
Total Moment 0.26 Mo 0.52 Mo 0.70 Mo 0.35 Mo 0.65 Mo
Column Strip 0.26 Mo 0.31 Mo 0.53 Mo 0.21 Mo 0.49 Mo
Middle Strip 0 0.21 Mo 0.17 Mo 0.14 Mo 0.16 Mo
Note: All negative moments are at face of support.
21 3 4 5
Table 4-2 Flat Plate or Flat Slab Supported Directly on Columns

4-8
Slab Moments
1
Exterior
Negative
2
Positive
3
First Interior
Negative
4
Positive
5
Interior
Negative
Middle Strip 0.16 Mo 0.14 Mo 0.16 Mo 0.14 Mo 0.16 Mo
Table 4-4 Flat Plate or Flat Slab with End Span Integral with Wall
1 2 3 4 5
Slab Moments
1
Exterior
Negative
2
Positive
3
First Interior
Negative
4
Positive
5
Interior
Negative
Note: (1) All negative moments are at face of support.
(2) Torsional stiffness of spandrel beams βt βt
βt
> 2.5. For values of less than 2.5,
exterior negative column strip moment increases to (0.30 – 0.03 ) Mo.
Table 4-3 Flat Plate or Flat Slab with Spandrel Beams
1 2 3 4 5

4-9
Span
ratio
Slab Moments
1
Exterior
Negative
2
Positive
3
First Interior
Negative
4
Positive
5
Interior
Negative
12 / Total Moment 0.16 Mo 0.57 Mo 0.70 Mo 0.35 Mo 0.65 Mo
Column Strip Beam 0.12 Mo 0.43 Mo 0.54 Mo 0.27 Mo 0.50 Mo
0.5 Slab 0.02 Mo 0.08 Mo 0.09 Mo 0.05 Mo 0.09 Mo
1.0 Slab 0.02 Mo 0.06 Mo 0.08 Mo 0.04 Mo 0.07 Mo
2.0 Slab 0.01 Mo 0.04 Mo 0.05 Mo 0.02 Mo 0.04 Mo
Note: (1) Beams and slab satisfy stiffness criteria: 121 / 1.0 and t 2.5.
(2) Interpolated between values shown for different ratios.12 /
(3) All negative moments are at face of support.
(4) Concentrated loads applied directly to beams must be accounted for separately.
Slab Moments
1
Exterior
Negative
2
Positive
3
First Interior
Negative
4
Positive
5
Interior
Negative
Total Moment 0 0.63 Mo 0.75 Mo 0.35 Mo 0.65 Mo
Column Strip 0 0.38 Mo 0.56 Mo 0.21 Mo 0.49 Mo
Middle Strip 0 0.25 Mo 0.19 Mo 0.14 Mo 0.16 Mo
Table 4-5 Flat Plate or Flat Slab with End Span Simply Supported on Wall
1 2 3 4 5
Table 4-6 Two-Way Beam-Supported Slab
1 2 3 4 5
αf1
˜2
/˜1
≥ 1.0 and βt
≥ 2.5

4-10
CL CL
CL
Slab, Is
Slab, Is
Beam,
Ib
Beam, Ib
b
h
h
b
a
a
b + 2 (a - h) ≤ b + 8h
b + (a - h) ≤ b + 4h
˜2/2 + c1/2
˜2
c1
Figure 4-6 Effective Beam and Slab Sections for Stiffness Ratio αf
(ACI 13.2.4)
The moment coefficients of Table 4-3 (flat plate with spandrel beams) are valid for βt
≥ 2.5, the coefficients of
Table 4-6 (two-way beam-supported slabs), are applicable when αf1
˜2
/˜1
≥ 1.0 and βt
≥ 2.5 (βt
, and αf1
are
stiffness parameters defined below). Many practical beam sizes will provide beam-to-slab stiffness ratios such
that αf1
˜2
/˜1
and βt
would be greater than these limits, allowing moment coefficients to be taken directly from
the tables. However, if beams are present, the two stiffness parameters αf1
and βt
will need to be evaluated.
For two-way slabs, the stiffness parameter αf1
is simply the ratio of the moments of inertia of the effective beam
and slab sections in the direction of analysis, αf1
= Ib
/Is
, as illustrated in Fig. 4-6. Figures 4-7 and 4-8 can be
used to determine αf
.
Relative stiffness provided by a spandrel beam is reflected by the parameter βt
= C/2Is
, where Is
is the moment
of inertia of the effective slab spanning in the direction of ˜1
and having a width equal to ˜2
, i.e., Is
= ˜2
h3
/12.
The constant C pertains to the torsional stiffness of the effective spandrel beam cross section. It is found
by dividing the beam section into its component rectangles, each having smaller dimension x and larger
dimension y, and summing the contribution of all the parts by means of the equation.
The subdivision can be done in such a way as to maximize C. Figure 4-9 can be used to determine the torsional
constant C.
C = 1− 0.63
x
y
⎛
⎝
⎜
⎞
⎠
⎟∑
x3
y
3

2.8
2.7
2.6
2.5
2.4
2.3
2.2
2.1
2.o
1.9
1.8
1.7
1.6
1.5
1.4
1.3
1.2
1.1
1.0
1 1.5 2 3 4 5 6 7 8 9 10
a/h
f 1.6
5
10
20
1.8
7
b/h = 0.4
0.5
0.6
0.8
b/h = 1
1.2
1.4
2
2.4
3
b/h = 4
4-11
h
a
b
b + 2 (a - h) ≤ b + 8h
CLCL
αf =
b
2
a
h
3
f
˜
2˜
Figure 4-7 Beam to Slab Stiffness Ratio αf (Interior Beam)

a/h
f
2.5
2.4
2.3
2.2
2.1
2.0
1.9
1.8
1.7
1.6
1.5
1.4
1.3
1.2
1.1
1.0 1.5 2 3 4 5 6 7 8 9 10
2.4
b/h = 0.4
0.6
0.8
b/h = 1.0
1.6
2.0
1.2
1.4
5.0
10.0
20.0
b/h = 3.0
4-12
a
b
˜2/2 + b/2
CL
b + (a – h) ≤ b + 4h
h
αf =
b a
h
3
f
2˜
Figure 4-8 Beam to Slab Stiffness Ratio αf (Spandrel Beams)

x*
y
12
14
16
18
20
22
24
27
30
33
36
42
48
54
60
4
202
245
288
330
373
416
458
522
586
650
714
842
970
1,098
1,226
5
369
452
535
619
702
785
869
994
1,119
1,244
1,369
1,619
1,869
2,119
2,369
6
592
736
880
1,024
1,168
1,312
1,456
1,672
1,888
2,104
2,320
2,752
3,184
3,616
4,048
7
868
1,096
1,325
1,554
1,782
2,011
2,240
2,583
2,926
3,269
3,612
4,298
4,984
5,670
6,356
8
1,188
1,529
1,871
2,212
2,553
2,895
3,236
3,748
4,260
4,772
5,284
6,308
7,332
8,356
9,380
9
1,538
2,024
2,510
2,996
3,482
3,968
4,454
5,183
5,912
6,641
7,370
8,828
10,286
11,744
13,202
10
1,900
2,567
3,233
3,900
4,567
5,233
5,900
6,900
7,900
8,900
9,900
11,900
13,900
15,900
17,900
12
2,557
3,709
4,861
6,013
7,165
8,317
9,469
11,197
12,925
14,653
16,381
19,837
23,293
26,749
30,205
14
—
4,738
6,567
8,397
10,226
12,055
13,885
16,629
19,373
22,117
24,861
30,349
35,837
41,325
46,813
16
—
—
8,083
10,813
13,544
16,275
19,005
23,101
27,197
31,293
35,389
43,581
51,773
59,965
68,157
*Small side of a rectangular cross section with dimensions x and y.
Values of torsion constant, C = (1– 0.63 x/y)(x3 y/3)
4-13
hb
a
a− h ≤ 4h
x2
x1
y1
y2
(1)
x1
y1
y2
(2)
Spandrel beam (ACI 13.2.4) Use larger value of C computed from (1) or (2)
x2
Figure 4-9 Design Aid for Computing Torsional Section Constant C

4-14
The column strip and middle strip moments are distributed over an effective slab width as illustrated in
Fig. 4-10. The column strip is defined as having a width equal to one-half the transverse or longitudinal span;
whichever is smaller (ACI 13.2.1). The middle strip is bounded by two column strips.
Middlestrip
1/2middlestrip
Interiorcolumnstrip
Exteriorcolumnstrip
˜2˜2
˜2/2 ˜2/2 ˜2/2
˜2/4 ˜2/4 ˜2/4
˜1
(a) Column strip for ˜2 < ˜1
Middlestrip
1/2middlestrip
Interiorcolumnstrip
Exteriorcolumnstrip
˜2˜2
˜2/2 ˜2/2 ˜2/2
˜1/4 ˜1/4 ˜1/4
˜1
(b) Column strip for ˜2 > ˜1
Figure 4-10 Definition of Design Strips
Once the slab and beam (if any) moments are determined, design of the slab and beam sections follows
the simplified design approach presented in Chapter 3. Slab reinforcement must not be less than that given in
Table 3-5, with a maximum spacing of 2h or 18 in. (ACI 13.3).

4.4 SHEAR IN TWO-WAY SLAB SYSTEMS
When two-way slab systems are supported by beams or walls, the shear capacity of the slab is seldom a critical
factor in design, as the shear force due to the factored loads is generally well below the capacity of the concrete.
In contrast, when two-way slabs are supported directly by columns (as in flat plates and flat slabs), shear
near the columns is of critical importance. Shear strength at an exterior slab-column connection (without
spandrel beams) is especially critical because the total exterior negative slab moment must be transferred
directly to the column. This aspect of two-way slab design should not be taken lightly by the designer.
Two-way slab systems will normally be found to be fairly “forgiving” if an error in the distribution or even
in the amount of flexural reinforcement is made, but there will be no forgiveness if a critical lapse occurs in
providing the required shear strength.
For slab systems supported directly by columns, it is advisable at an early stage in the design to check the shear
strength of the slab in the vicinity of columns as illustrated in Fig. 4-11.
4-15
Edge
column
C panelL
Interior
column
Corner
column
Critical shear
perimeter
Effective area for direct
shear force
Figure 4-11 Critical Locations for Slab Shear Strength

4.4.1 Shear in Flat Plate and Flat Slab Floor Systems
Two types of shear need to be considered in the design of flat plates or flat slabs supported directly on columns.
The first is the familiar one-way or beam-type shear, which may be critical in long narrow slabs. Analysis for
beam shear considers the slab to act as a wide beam spanning between the columns. The critical section is
taken a distance d from the face of the column. Design against beam shear consists of checking the
requirement indicated in Fig. 4-12(a). Beam shear in slabs is seldom a critical factor in design, as the shear
force is usually well below the shear capacity of the concrete.
Two-way or “punching” shear is generally the more critical of the two types of shear in slab systems supported
directly on columns. Punching shear considers failure along the surface of a truncated cone or inverted pyramid
around a column. The critical section is taken perpendicular to the slab at a distance d/2 from the perimeter of
a column. The shear force Vu
to be resisted can be easily calculated as the total factored load on the area
bounded by panel centerlines around the column, less the load applied within the area defined by the critical
shear perimeter (see Fig. 4-11). In the absence of a significant moment transfer from the slab to the column,
design against punching shear consists of ensuring that the requirement in Fig. 4-12(b) is satisfied. Figures 4-13
through 4-15 can be used to determine φVc
for interior, edge and corner columns, respectively.
4-16
CL panels
(a) Beam shear
Critical
section
d
˜1
˜2
CL panels˜1
˜2
d/2
Critical
section
(b) Two-way shear
Figure 4-12 Direct Shear at an Interior Slab-Column Support (for normal weight concrete and › = 4000 psi)
where Vu
is factored shear force (total factored load on
shaded area). Vu
is in kips and ˜2
and d are in inches.
Vu
= factored shear force (total factored load on shaded area), kips
bo
= perimeter of critical section, in.
β = long side/short side of reaction area
αs
= constant (ACI 11.11.2.1 (b))
Vu
≤ φVc
where :
φVc = least of
φ 2 +
4
β
⎛
⎝
⎜
⎞
⎠
⎟ ʹfc
bo
d = 0.048 2 +
4
βc
⎛
⎝
⎜
⎞
⎠
⎟ bo
d
φ
αs
d
bo
+ 2
⎛
⎝
⎜
⎞
⎠
⎟ ʹfc
bo
d = 0.048
αs
d
bo
+ 2
⎛
⎝
⎜
⎞
⎠
⎟ bo
d
φ4 ʹfc
bo
d = 0.19bo
d
⎧
⎨
⎪
⎪
⎪
⎪
⎪
⎩
⎪
⎪
⎪
⎪
⎪
Vu
≤ φVc
≤ φ2 ʹfc 2 d
≤ 0.095 2 d ( ʹfc = 4000psi)

4-17
0.1
0.11
0.12
0.13
0.14
0.15
0.16
0.17
0.18
0.19
0.2
5 15 25 35 45 55 65 75 85 95
bo/d
4
5
6
7
9
10
15
20
φVc/bod,ksi
1β= , 2
3
8
Figure 4-13 Two-Way Shear Strength of Slabs, Interior Column (αs = 40)

4-18
0.1
0.11
0.12
0.13
0.14
0.15
0.16
0.17
0.18
0.19
0.2
5 15 25 35 45 55 65 75 85 95
bo/d
φVc/bod,ksi
4
5
6
7
9
10
15
20
β =1, 2
3
8
Figure 4-14 Two-Way Shear Strength of Slabs, Edge Column (αs = 30)

4-19
0.1
0.11
0.12
0.13
0.14
0.15
0.16
0.17
0.18
0.19
0.2
5 15 25 35 45 55 65 75 85 95
bo/d
4
5
6
7
9
10
15
20
β=1, 2
φVcbod,ksi
3
8
Figure 4-15 Two-Way Shear Strength of Slabs, Corner Column (αs = 20)

For practical design, only direct shear (uniformly distributed around the perimeter bo
) occurs around interior
slab-column supports where no (or insignificant) moment is to be transferred from the slab to the column.
Significant moments may have to be carried when unbalanced gravity loads on either side of an interior column
or horizontal loading due to wind must be transferred from the slab to the column. At exterior slab-column
supports, the total exterior slab moment from gravity loads (plus any wind moments) must be transferred
directly to the column.
Transfer of unbalanced moment between a slab and a column takes place by a combination of flexure (ACI
13.5.3) and eccentricity of shear (ACI 11.11.7). Shear due to moment transfer is assumed to act on a critical
section at a distance d/2 from the face of the column, the same critical section around the column as that used
for direct shear transfer [Fig. 4-12(b)]. The portion of the moment transferred by flexure is assumed to be
transferred over a width of slab equal to the transverse column width c2
, plus 1.5 times the slab or drop panel
thickness (1.5h) on each side of the column. Concentration of negative reinforcement is to be used to resist
moment on this effective slab width. The combined shear stress due to direct shear and moment transfer often
governs the design, especially at the exterior slab-columns supports.
The portions of the total moment to be transferred by eccentricity of shear and by flexure are given by ACI Eqs.
(11-37) and (13-1), respectively. For square interior or corner columns, 40% of the moments is considered
transferred by eccentricity of the shear (γ
v
Mu
= 0.40 Mu
), and 60% by flexure (γ
f
Mu
= 0.60 Mu
), where Mu
is
the transfer moment at the centroid of the critical section. The moment Mu
at an exterior slab-column
support will generally not be computed at the centroid of the critical transfer section in the frame analysis.
In the Direct Design Method, moments are computed at the face of the support. Considering the approximate
nature of the procedure used to evaluate the stress distribution due to moment transfer, it seems unwarranted to
consider a change in moment to the transfer centroid; use of the moment values at the faces of the supports
would usually be accurate enough.
The factored shear stress on the critical transfer section is the sum of the direct shear and the shear caused by
moment transfer,
vu
= Vu
/Ac
+ γ
v
Mu
c/J
or
vu
= Vu
/Ac
– γ
v
Mu
c' /J
Computation of the combined shear stress involves the following properties of the critical transfer section:
Ac
= area of critical section, in.2
c or c' = distance from centroid of critical section to the face of section where stress
is being computed, in.
Jc
= property of critical section analogous to polar moment of inertia, in.4
The above properties are given in terms of formulas in Tables 4-7 through 4-10 (located at the end of this
chapter) for the four cases that can arise with a rectangular column section: interior column (Table 4-7), edge
column with bending parallel to the edge (Table 4-8), edge column with bending perpendicular to the edge
(Table 4-9), and corner column (Table 4-10). Numerical values of the above parameters for various
combinations of square column sizes and slab thicknesses are also given in these tables. Properties of the
4-20

critical shear transfer section for circular interior columns can be found in Reference 4.2. Note that in the case
of flat slabs, two different critical sections need to be considered in punching shear calculations as shown in
Fig. 4-16. Tables 4-7 through 4-10 can be used in both cases. Also, Fig. 4-17 can be used to determine γ
v
and
γ
f
given b1
and b2
.
Unbalanced moment transfer between slab and an edge column (without spandrel beams) requires special
consideration when slabs are analyzed by the Direct Design Method for gravity loads. To assure adequate shear
strength when using the approximate end-span moment coefficient, the moment 0.30 Mo
must be used in
determining the fraction of unbalanced moment transferred by eccentricity of shear (γ
v
Mu
= γ
v
0.30Mo
)
according to ACI 13.6.3.6. For end spans without spandrel beams, the column strip is proportioned to resist the
total exterior negative factored moment (Table 4-2). The above requirement is illustrated in Fig. 4-18. The total
reinforcement provided in the column strip includes the additional reinforcement concentrated over the column
to resist the fraction of unbalanced moment transferred by flexure γ
f
Mu
= γ
f
(0.26Mo
), where the moment
coefficient (0.26) is from Table 4-2. Application of this special design requirement is illustrated in Section 4.7.
4.5 COLUMN MOMENTS DUE TO GRAVITY LOADS
Supporting columns (and walls) must resist any negative moments transferred from the slab system.
For interior columns, the approximate ACI Eq. (13-7) may be used for unbalanced moment transfer due to
gravity loading, unless an analysis is made considering the effects of pattern loading and unequal adjacent
spans. The transfer moment is computed directly as a function of the span length and gravity loading. For the
more usual case with equal transverse and longitudinal spans, ACI Eq. (13-7) simplifies to:
Mu
= 0.07(0.05qLu
˜2
ñ
2) = 0.035qLu
˜2
ñ
2
where qLu
= factored live load, psf
˜2
= span length transverse to
ñ
= clear span length in direction Mu
is being determined
4-21
d2/2 d1/2
d1
d2
Critical
sections
Figure 4-16 Critical Shear-Transfer Sections for Flat Slabs

4-22
b2/b1
b1/b2
0.0 0.2 0.4 0.6 0.8 1.0
0.0 0.2 0.4 0.6 0.8 1.0
100
80
60
40
20
0
0
20
40
60
80
100
γf–percentbyflexure
γv–percentbyshear
for b1 ≥ b2
for b1 ≤ b2
Figure 4-17 Solution to ACI Equations (11-37) and (13-1)
h
γv
Mu
=γv
0.3Mo
0.26Mo
*
γ f
(0.26Mo
)Effective slab width
for moment
transfer by flexure
c +2 (1.5h)
c
*Total Moment
Colum
n
Strip
Figure 4-18 Unbalanced Moment Transfere at Edge Column (ACI 13.6.3.6)

At an exterior column, the total exterior negative moment from the slab system is transferred directly to the
column. Due to the approximate nature of the moment coefficients of the Direct Design Method, it seems
unwarranted to consider the change in moment from face of support to centerline of support; use of the
exterior negative slab moment directly would usually be accurate enough.
Columns above and below the slab must resist portions of the support moment based on the relative
column stiffnesses (generally, in proportion to column lengths above and below the slab). Again, due to the
approximate nature of the moment coefficients, the refinement of considering the change in moment from
centerline of slab to top or bottom of slab seems unwarranted.
4.6 REINFORCEMENT DETAILING
In computing required steel areas and selecting bar sizes, the following will ensure conformance to the Code
and a practical design.
1. Minimum reinforcement area = 0.0018 bh (b = slab width, h = total thickness) for Grade 60 bars for either
top or bottom steel. These minima apply separately in each direction (ACI 13.3.1).
2. Maximum bar spacing is 2h, but not more than 18 in. (ACI 13.3.2).
3. Maximum top bar spacing at all interior locations subject to construction traffic should be limited. Not less
than No. 4 @ 12 in. is recommended to provide adequate rigidity and to avoid displacement of top bars with
standard bar support layouts under ordinary foot traffic.
4. Maximum ρ = As
/bd is limited to 0.0206 however, ρ ≤ 0.50 ρmax
is recommended to provide deformability,
to avoid overly flexible systems subject to objectionable vibration or deflection, and for a practical balance
to achieve overall economy of materials, construction and design time.
5. Generally, the largest size of bars that will satisfy the maximum limits on spacing will provide overall
economy. Critical dimensions that limit size are the thickness of slab available for hooks and the distances
from the critical design sections to edges of slab.
4.7 EXAMPLES: SIMPLIFIED DESIGN FOR TWO-WAY SLABS
The following two examples illustrate the use of the simplified design data presented in this chapter for the
analysis and design of two-way slab systems. The two-way slab system for Building #2 is used to illustrate the
simplified design.
4-23

4.7.1 Example: Interior Strip (N-S Direction) of Building #2, Alternate (2)
The slab and column framing will be designed for gravity loads only; the structural walls will carry the total
wind forces.
1. Data: › = 4000 psi (carbonate aggregate)
fy
= 60,000 psi
Floors: LL = 50 psf
DL = 142 psf (9 in. slab + 20 psf partitions
+ 10 psf ceiling and misc.)
Required fire resistance rating = 2 hours
Preliminary slab thickness:
Determine preliminary h based on two-way shear at an interior
column (see Fig. 1-8).
Live load reduction: Influence area AI (4 panels) = 24 x 20 x 4 = 1920 sq ft
L = 50(0.25 + 15/ ) = 29.5 psf
qu
= 1.2(142) + 1.6(29.5) = 218 psf
A = 24 ϫ 20 = 480 ft2
c1
2
= 16 ϫ 16 = 256 in.2
= 1.8 ft2
A/c1
2
= 480/1.8 = 267
From Fig. 1-8, required d/c1
ഡ 0.38
Required d = 0.38 ϫ 16 = 6.1 in.
h = 6.10 + 1.25 = 7.35 in.
Determine preliminary h based on two-way shear at an exterior column (see Fig. 1-9).
Live load reduction: AI (2 panels) = (24 ϫ 2 ϫ 20.5 = 984 sq ft)
L = 50(0.25 + 15/984 ) = 36.4 psf
qu = 1.2(142) + 1.6(36.4) = 229 psf
A = 24(20/2 + 0.5) = 252 ft2
c1
2 = 12 ϫ 12 = 144 in.2
= 1.0 ft2
A/c1
2 = 252/1.0 = 252
From Fig. 1-9, required d/c1
= 0.52
Required d = 0.52 ϫ 12 = 6.2 in.
h = 6.2 + 1.25 = 7.45 in. governs
1920
4-24
Design
strip
3@20'-0"
24'-0" 24'-0"
12" x 12"
16" x 16"

Determine preliminary h based on two-way shear at an corner column (see Fig. 1-10).
Live load reduction: AI (one panels) = 24.5 ϫ 20.5 = 502 sq ft
L = 50(0.25 + 15/502) = 46.0 psf
qu
= 1.2(142) + 1.6(46.0) = 244 psf
A = (20/2 + 0.5) (24/2 + 0.5) = 131.25 ft2
c1
2 = 12 ϫϾ 12 = 144 in.2 = 1.0 ft2
A/c1
2
= 131.25/1.0 = 131.25
From Fig. 1-10, required d/c1 = 0.47
Required d = 0.47 ϫ 12 = 5.6 in.
h = 5.6 + 1.25 = 6.85 in.
To account for moment transfer at the edge columns, increase h by 20%
Try preliminary h = 9 in.
2. Check the preliminary slab thickness for deflection control and shear strength.
(a) Deflection control:
From Table 4-1 (flat plate): h = ˜n
/30 = (22.67 ϫ 12)/30 = 9.07 in.
where ˜n
= 24 – (16/12) = 22.67 ft
(b) Shear Strength:
From Fig. 4-13: Check two-way shear strength at interior slab column support for h = 9 in.
From Table 4-7: Ac
= 736.3 in.2
for 9 in. slab with 16 ϫ 16 in column.
b1
= b2
= 2(11.88) = 23.76 in. = 1.98 ft
Vu
= 0.218(24 ϫ 20 – 1.98*) = 103.8 kips
From Fig. 4-13, with β = 1 and bo
/d = 4(23.76)/(9 – 1.25) = 12.3:
φVc
= 0.19 Ac
= 0.19(736.3) = 139.9 kips > 118.5 kips O.K.
Check for fire resistance: From Table 10-1, for fire resistance rating of 2 hours, required slab thickness =
4.6 in. < 9.0 in. O.K.
Use 9 in. slab.
4-25

3. Check limitations for slab analysis by coefficients (ACI 13.6.1)
• 3 continuous spans in one direction, 5 in the other
• rectangular panels with long-to-short span ratio = 24/20 = 1.2 < 2
• successive span lengths in each direction are equal
• LL/DL = 50/142 = 0.35 < 2
• slab system is without beams
Since all requirements are satisfied, the Direct Design Method can be used to determine the moments.
4. Factored moments in slab (N-S direction)
(a) Total panel moment Mo
:
Mo
= qu
˜2
ñ
2
/8
= 0.245 x 24 x 18.832
/8 = 260.6 ft-kips
where qu
= 1.2(142) + 1.6(46.5*) = 245 psf
˜2
= 24 ft
ñ
(interior span) = 20 – 1.33 = 18.67 ft
ñ
(end span) = 20 – 0.67 – 0.50 = 18.83 ft
Use larger value of ñ
for both spans.
(b) Negative and positive factored moments:
into negative and positive moments, and then, column and
middle strip moments, involves direct application of the moment coefficients in Table 4-2.
4-26
End Spans Interior Span
Slab Moments
(ft-kips)
Exterior
Negative Positive
Interior
Negative Positive
Total Moment 67.8 135.5 182.4 91.2
Column Strip 67.8 80.8 138.1 54.7
Middle Strip 0 54.7 44.3 36.5
Note: All negative moments are at face of columns.
* Live load reduction: AI
(1 panel) = 24 x 20 = 480 sq ft
L = 50(0.25 + 15/ ) = 46.5 psf480

5. Slab Reinforcement
Required slab reinforcement is easily determined using a tabular form as follows:
6. Check slab reinforcement at exterior column (12 ϫ 12 in.) for moment transfer between slab and column.
For a slab without spandrel beams, the total exterior negative slab moment is resisted by the column strip
(i.e., Mu
= 67.8 ft-kips).
Fraction transferred by flexure using ACI Eq. (13-1):
b1
= 12 + (7.75/2) = 15.88 in.
b2
= 12 + 7.75 = 19.75 in.
From Fig. 4-17, γ f ഡ 0.62 with b1/b2 = 0.8
Mu = 0.62 (67.8) = 42.0 ft-kips
As
= Mu
/4d = 42/(4 ϫ 7.75) = 1.35 in.2
No. of No. 4 bars = 1.35/0.20 = 6.75 bars, say 7 bars
Must provide 7-No. 4 bars within an effective slab width (ACI 13.5.3.3) = 3h + c2
= 3(9) + 12 = 39 in.
Provide the required 7-No. 4 bars by concentrating 7 of the column strip bars (13-No. 4) within the
3 ft-3 in. slab width over the column. For symmetry, add one column strip bar to the remaining 5 bars so
that 3 bars will be on each side of the 3 ft-3 in. strip. Check bar spacing:
For 7-No. 4 within 39 in. width: 39.8 = 4.9 in.
For 6-No. 4 within (120 – 39) = 81 in. width: 81/6 = 13.5 in. < 18 in. O.K.
No additional bars are required for moment transfer.
4-27
Span Location Mu
(ft-kips)
b1
(in.)
b2
(in.)
As =
Mu/4d
(in. 2
)
As
3
(min)
(in. 2
)
No. of
#4 Bars
No. of
#5 Bars
END SPAN
Column Ext. Negative 67.8 120 7.75 2.19 1.94 11 8
Strip Positive 80.8 120 7.75 2.61 1.94 14 9
Int. Negative 138.1 120 7.75 4.45 1.94 23 15
Middle Ext. Negative 0 168 7.75 0.00 2.72 14 9
Strip Positive 54.7 168 7.75 1.76 2.72 14 9
Int. Negative 44.3 168 7.75 1.43 2.72 14 9
INTERIOR SPAN
Column Positive 54.7 120 7.75 1.76 1.94 10 7
Strip
Middle Positive 36.5 168 7.75 1.18 2.72 14 9
Strip
1
Column strip = 0.5(20 x 12) = 120 in. (see Fig. 4-9b)
Middle Strip = (24 x 12) – 120 = 168 in.
2
Use average d = 9 – 1.25 = 7.75 in.
3
As(min) = 0.0018 bh = 0.0162b
smax = 2h < 18 in = 2(9) = 18 in.

4-28
7. Reinforcement details are shown in Figs. 4-19, 4-20, and 4-22. Bar lengths are determined directly from
Fig. 8-6. Note that for structural integrity, all bottom bars within the column strip, in each direction must be
continuous or spliced with class B splices. Splices must be located as shown in the figure. At least two of
the column strip bottom bars must pass within the column core and must be anchored at exterior supports.
Column Strip
14-#414-#4
7-#4
7-#4
7-#4
7-#4
4'-10" 4'-10"4'-2"
6"
0"
3'-0" 3'-0"
6"
Middle Strip
13-#4
7-#4
7-#4
20'-0" 20'-0"
15-#5
6'-4" 6'-4"
6"
12"
5'-8"
1'-4"
Splices are permitted
in this region
Figure 4-19 Reinforcement Details for Flat Plate of Building #2—
Alternate (2) Interior Slab Panel (N-S Direction)

4-29
8. Check slab shear strength at edge column for gravity load shear and moment transfer (see Fig. 4-21).
(a) Direct shear from gravity loads:
live load reduction:
KLL
AT
(2 panels) = 24 ϫ 20 ϫ 2 = 960 sq ft
L = 50(0.25 + 15/ ) = 36.7 psf
qu
= 1.2(142) + 1.6(36.7) = 229 psf
Vu
= 0.229[(24)(10.5) – (1.32)(1.65)] = 57.2 kips
(b) Moment transfer from gravity loads:
When slab moments are determined using the approximate
moment coefficients, the special provisions of ACI 13.6.3.6
apply for moment transfer between slab and an edge column.
The fraction of unbalanced moment transferred by eccentricity
of shear (ACI 11.11.7.1) must be based on 0.3 Mo.
(c) Combined shear stress at inside face of critical transfer
section:
960
12" x 12"
column
Column strip—10'-0"
(13-#4T)
7-#4T
3'-3"
Figure 4-20 Bar Layout Detail for 13-#4 Top Bars at Exterior Columns
24.0'
Slab
edge
9 in. slab (d = 7.75 in.)
10.5'
Critical section
1.65'
1.32'
Figure 4-21 Critical Section
for Edge Column

4-30
From Table 4-9, for 9 in. slab with 12 ϫ 12 in. column:
Ac = 399.1 in.2
J/c = 2523 in.3
From Fig. 4-17, with b1/b2 = 15.88/19.75 = 0.8, γv ഡ 0.38
vu
= Vu
/Ac
+ γ
v
M c/J
= (57,200/399.1) + (0.38 ϫ 0.3 ϫ 260.6 ϫ 12,000/2523)
= 143.3 + 139.4 = 284.6 psi >> = 190 psi*φ4 ʹfc
Column
CL
Column
CL
ColumnCL
ColumnCL
Column strip Column stripMiddle strip
20'-0"20'-0"
Slab edge5'-0" 5'-0" 5'-0" 5'-0"
13-#4T* 13-#4T*14-#4T
14-#4B14-#4B 14-#4B 14-#4B
15-#5T 14-#4 15-#5T
14-#4B 10-#4B10-#4B
24'-0"
*See special bar layout detail in Fig. 4-19
Figure 4-22 Bar Layout—Space Bars Uniformly within Each Column Strip and Middle Strip
* bo/d = [(2 x 15.88) + 19.75]/7.75 = 6.7; from Fig. 4-13 with βc = 1, φVc/bod = 190 psi

4-31
The 9 in. slab is not adequate for shear and unbalanced moment transfer at the edge columns. Increase
shear strength by providing drop panels at edge columns. Calculations not shown here.
4.7.2 Example: Interior Strip (N-S Direction) of Building #2, Alternate (1)
The slab and column framing will be designed for both gravity and wind loads. Design an interior strip for the
1st-floor level (greatest wind load effects).
1. Data: › = 4000 psi (carbonate aggregate)
fy = 60,000 psi
Floors: LL = 50 psf
DL = 136 psf (81
⁄2 in. slab + 20 psf partitions
+ 10 psf ceiling & misc.)
Preliminary sizing: Slab = 81
⁄2 in. thick
Exterior = 12 ϫ 12 in.
Spandrel beams = 12 ϫ 20 in.
2. Determine the slab thickness for deflection control and shear strength
(a) Deflection control:
From Table 4-1 (flat plate with spandrel beams, αf ≥ 0.8):
h = ˜n/33 = (22.67 ϫ 12)/33 = 8.24 in.
where ˜n = 24 – (16/12) = 22.67 ft
(b) Shear strength. Check shear strength for an 81
⁄2 in. slab:
With the slab and column framing designed for both gravity and wind loads, slab shear strength needs
to be checked for the combination of direct shear from gravity loads plus moment transfer from wind
loads. Wind load analysis for Building #2 is summarized in Fig. 2-15. Moment transfer between slab
and column is greatest at the 1st-floor level where wind moment is the largest. Transfer moment (unfac-
tored) at 1st-floor level due to wind, Mw
= 55.78 + 55.78 = 111.56 ft-kips.
Design
strip
12" x 20" spandrel beam
3@20'-0"
24'-0" 24'-0"
12" x 12"
16" x 16"

4-32
Direct shear from gravity loads:
qu
= 1.20(136) + 1.60(29.5*) = 210.4 psf
Vu
= 0.210(24 x 20 – 1.942) = 100.2 kips
where d = 8.50 – 1.25 = 7.25 in.
and b1
= b2
= (16 + 7.25)/12 = 1.94 ft
Gravity + wind load:
Two load combinations have to be considered ACI Eq. (9-3)
and Eq. (9-4).
ACI Eq. (9-3) [1.2D + 0.8W or 1.2D + 0.5L]
Vu
= [1.2(136) + 0.5(29.5)] x [24 x 20 – 1.942
] = 84.7 kips
Mu
= 0.8 ϫ 111.56 = 89.25 ft-kips
ACI Eq. (9-4) [1.2D + 0.5L + 1.6W]
Vu
= [1.2(136) + 0.5(29.5)] ϫ [24 ϫ 20 – 1.942
] = 84.75 kips
Mu
= 1.6 ϫ 111.56 = 178.5 ft-kips
From Table 4-7, for 81
⁄2 in. slab with 16 x 16 in. columns:
Ac
= 674.3 in.2
J/c = 5352 in.3
Shear stress at critical transfer section:
vu
= Vu
/Ac
+ γv
Mu
c/J
= (84,700/674.3) + (0.4 x 178.5 ϫ 12,000/5352)
= 125.6 + 160.1 = 285.7 psi > = 215 psi
The 81
⁄2 in. slab is not adequate for gravity plus wind load transfer at the interior columns.
Increase shear strength by providing drop panels at interior columns. Minimum slab thickness at drop
panel = 1.25(5.5) = 10.63 in. (see Fig. 4-2). Dimension drop to actual lumber dimensions for economy of
formwork. Try 21
⁄4 in. drop (see Table 9-1).
h = 8.5 + 2.25 = 10.75 in. > 10.63 in.
d = 7.25 + 2.25 = 9.5 in.
φ4 ʹfc
20'-0"
24'-0"
1.94'
1.94'
Critical
section
(4 panels) = 24 x 20 x 4 = 1920 sq ft
L = 50(0.25 + 15/ ) = 29.5 psf1920

4-33
Refer to Table 4-7:
b1
= b2
= 16 + 9.5 = 25.5 in. = 2.13 ft
Ac
= 4(25.5) ϫ 9.5 = 969 in.2
J/c = [25.5 ϫ 9.5(25.5 ϫ 4) + 9.53]/3 = 8522 in.3
Vu
= (84,700/969) + (0.4 ϫ 178.5 ϫ 12,000/8522)
= 87.4 + 100.5 = 187.9 psi < 215 psi
Note that the shear stress around the drop panel is much less than the allowable stress (calculations not
shown here).
With drop panels, a lesser slab thickness for deflection control is permitted. From Table 4-1 (flat slab with
spandrel beams): h = ˜n
/36 = (22.67 ϫ 12)/36 = 7.56 in. could possibly reduce slab thickness from 81
⁄2 to 8
in.; however, shear strength may not be adequate with the lesser slab thickness. For this example hold the
slab thickness at 81
⁄2 in. Note that the drop panels may not be required in the upper stories where the transfer
moment due to wind become substantially less (see Fig. 2-15).
Use 81
⁄2 in. slab with 21
⁄4 in. drop panels at interior columns of 1st story floor slab. Drop panel dimensions
= ˜/3 = 24/3 = 8 ft. Use same dimension in both directions for economy of formwork.
Check for fire resistance: From Table 10-1 for fire resistance rating of 2 hours, required slab thickness
= 4.6 in. < 8.5 in. O.K.
3. Factored moments in slab due to gravity load (N-S direction).
(a) Evaluate spandrel beam-to-slab stiffness ratio α and β t
:
Referring to Fig. 4-7:
˜2
= (20 ϫ 12)/2 = 120 in.
a = 20 in.
b = 12 in.
h = 8.5 in.
a/h = 20/8.5 = 2.4
b/h = 12/8.5 = 1.4
f ഡ 1.37
Note that the original assumption that the minimum h = ˜n
/33 is O.K. since αf
> 0.8 (see Table 4-1).
βt
C
21s
=
8425
2 7,680( )
= 0.55 < 2.5
α =
b
2
a / h( )3
f =
12
120
24( )3
1.37( ) = 1.89 > 0.8f
23.5"
20"
12"
8.5"
12 + 11.5 < 12 + 4(8.5)

4-34
where Is
= (12 ϫ 12 + 6)(8.5)3/12 = 7,680 in.4
C = larger value computed for the spandrel beam section (see Fig. 4-8).
(b) Total panel moment Mo
:
Mo
= qu
˜2
ñ
2/8
where qu
= 1.2(136) + 1.6(46.5*) = 238 psf
˜2
= 24 ft
ñ
(interior span) = 20 – 1.33 = 18.67 ft
ñ
(end span) = 20 – 0.67 – 0.50 = 18.83 ft
Use larger value for both spans.
(c) Negative and positive factored gravity load moments:
into negative and positive moments, and the, column strip
and middle strip moments involves direct application of the moment coefficients of Table 4-3.
Note that the moment coefficients for the exterior negative column and middle strip moments need to
be modified for βt
less than 2.5. For βt
= 0.29:
Column strip moment = (0.30 – 0.03 ϫ 0.29)Mo
= 0.29Mo
Middle strip moment = 0.30Mo
– 0.29Mo
= 0.01Mo
x1 = 8.5
y1 = 23.5
C1 = 3714
x2 = 11.5
y2 = 12
C2 = 2410
x1 = 12
y1 = 20
C1 = 7165
x2 = 8.5
y2 = 11.5
C2 = 1260
⌺C = 3714 + 2410 = 6124 ⌺C = 7165 + 1260 = 8425 (governs)
End Spans Interior Span
Slab Moments
(ft-kips)
Exterior
Negative Positive
Interior
Negative Positive
Total Moment 75.9 126.5 177.2 88.6
Column Strip 73.4 75.9 134.1 53.2
Middle Strip 2.5 50.6 43.1 38.1
Note: All negative moments are at faces of columns.
(4 panels) = 24 x 20 = 480 sq ft
L = 50(0.25 + 15/ ) = 46.5 psf480

4-35
4. Check negative moment sections for combined gravity plus wind load moments
(a) Exterior Negative:
Consider wind load moments resisted by column strip as defined in Fig. 4-9. Column strip width
= 0.5(20 ϫ 12) = 120 in.
gravity loads only:
Mu
= 73.4 ft-kips
gravity + wind loads:
Mu
= ϫ 73.4 + 1.6 ϫ 55.78 = 146.75 ft-kips (govern)
ACI Eq. (9-4)
Also check for possible moment reversal due to wind moments:
Mu
= 0.9(42) ± 1.6(55.78) = 127.0 ft-kips, – 51.4 ft-kips (reversal)
ACI Eq. (9-6)
where qD
= 136 psf
Md
= 0.29(0.136 ϫ 24 ϫ 18.832
/8) = 42 ft-kips
(b) Interior Negative:
gravity loads only:
Mu
= 134.1 ft-kips
Mu
= ϫ 134.1 + 1.6 ϫ 55.78 = 194.3 (governs)
and Mu = 0.9(76.8) ± 1.6(55.78) = 158.4 ft-kips, – 20.1 ft-kips (reversal)
where Md = 42(0.53/0.29) = 76.8 ft-kips
5. Check slab section for moment strength
(a) At exterior negative support section:
b = 20 Mu/d2
= 20 ϫ 146.75/7.252
= 55.8 in. < 120 in. O.K.
where d = 8.5 – 1.25 = 7.25 in.
(b) At interior negative support section:
b = 20 x 194.3/9.502
= 43.1 in. < 120 in. O.K.
1.2 136( ) + 0.5 46.5( )
238
1.2 136( ) + 0.5 46.5( )
238

4-36
where d = 7.25 + 2.25 = 9.50 in.
6. Slab Reinforcement
Required slab reinforcement is easily determined using a tabular form as follows:
7. Check slab reinforcement at interior columns for moment transfer between slab and column. Shear strength
of slab already checked for direct shear and moment transfer in Step (2)(b). Transfer moment (unfactored)
at 1st-story due to wind, Mw
= 111.56 ft-kips.
Fraction transferred by flexure using ACI Eqs. (13-1) and (9-6):
Mu
= 0.60(1.6 ϫ 111.56) = 107.1 ft-kips
As
= Mu
/4d = 107.1/(4 ϫ 9.50) = 2.82 in.2
For No. 5 bars, 2.82/0.31 = 9.1 bars, say 10-No. 5 bars
Must provide 10-No. 5 bars within an effective slab width = 3h + c2
= 3(10.75) + 16 = 48.3 in.
Provide the required 10-No. 5 bars by concentrating 10 of the column strip bars (18-No. 5) within the 4-ft
slab width over the column. Distribute the other 8 column strip bars (4 on each side) in the remaining
column strip width. Check bar spacing:
48/9 spaces = ± 5.3 in.
(120 – 48)/7 spaces = ± 10.3 in. < 18 in. O.K.
Span Location Mu
(ft-kips)
b1
(in.)
b2
(in.)
As =
Mu/4d
(in. 2
)
As
3
(min)
(in. 2
)
No. of
#4 Bars
No. of
#5 Bars
END SPAN
Column
Strip
Ext. Negative 146.75
-51.4
120
120
7.25
7.25
5.06
1.77
1.84
0.00
26
9
17
9
Positive 75.9 120 7.25 2.62 1.84 14 9
Int. Negative 194.3
-20.1
120
120
9.5
7.25
5.11
0.69
2.32
0.00
26
9
17
9
Middle Ext. Negative 2.5 168 7.25 0.09 2.57 13 9
Strip Positive 50.6 168 7.25 1.74 2.57 13 9
Int. Negative 43.1 168 7.25 1.49 2.57 13 9
INTERIOR SPAN
Column Positive 53.2 120 7.25 1.83 1.84 10 9
Strip
Middle Positive 40 168 7.25 1.38 2.57 13 9
Strip
1
Column strip = 0.5(20 x 12) = 120 in. (see Fig. 4-9b)
Middle Strip = (24 x 12) – 120 = 168 in.
2
Use average d = 8.5 – 1.25 = 7.25 in.
At drop panel, d = 7.25 + 2.25 = 9.50 in. (negative moment only)
3
As(min) = 0.0018 bh
smax = 2h < 18 in = 2(8.5) = 17 in.

4-37
Reinforcement details for the interior slab are shown in Figs. 4-23 and 4-24. Bar lengths for the middle strip
are taken directly from Fig. 8-6. For the column strip, the bar lengths given in Fig. 8-6 (with drop panels)
need to be modified to account for wind moment effects. In lieu of a rigorous analysis to determine bar cut-
offs based on a combination of gravity plus wind moment variations, provide bar length details as follows:
For bars near the top face of the slab, cut off one-half of the bars at 0.2˜n
from supports and extend the
remaining half the full span length, with a Class B splice near the center of span. Referring to Table 8-1,
splice length = 1.3 ϫ 23.7 = 30.8 in. ഡ 2.5 ft. At the exterior columns, provide a 90° standard hook with
2 in. minimum cover to edge of slab. From Table 8-2, for No. 5 bars, ˜dh
= 9 in. < 12 – 2 = 10 in. O.K. (For
easier bar placement, alternate equal bar lengths at interior column supports.)
At the exterior columns, provide a 90° end-hook with 2 in. minimum cover to edge of slab for all bottom
bars. At least 2 of the bottom bars in the column strip must pass within the column core and anchored at the
exterior support.
2'-6"
Column Strip
Middle Strip
Spandrel beam
Drop
panel
*See bar layout detail in Fig. 4-24
8.5"
9-#5*
3-#40"
0"
0"
0"
0"
4'-6"4'-6"
4'-0" 4'-0"
20'-0"20'-0"
9-#5*
9-#59-#5
3'-10"
2.25"
16"
2'-6" 7"7" 7"7-#47-#4
1'-8"
2" min.
0"
12"
7-#4
7-#4 7-#4
13-#4
13-#4
6-#46-#4
3'-0" 3'-0"
4'-10" 4'-10"
3'-10"
1'-8"
6"
6"
Splices are permitted
in this region
Figurer 4-23 Reinforcement Details for Flat Slab of Building #2 Alternate (1)—1st Floor Interior Slab Panel
(N-S Direction)

4-38
4'-6"
4'-6"
4-#5T10 -#5T4-#5T
4'- 0"
Column strip (18-#5T)
3'- 0" 3'- 0"
Figure 4-24 Bar Layout Detail for 18-#5 Top Bars at Interior Columns

4-39
h = 5 in.,
d = 33/4 in.
h = 51/2 in.,
d = 41/4 in.
h = 6 in.,
d = 43/4 in.
h = 61/2 in.,
d = 51/4 in.
COL.
SIZE
Ac
in.2
J/c =
J/c'
in.3
c = c'
in.
Ac
in.2
J/c =
J/c'
in.3
c = c'
in.
Ac
in.2
J/c =
J/c'
in.3
c = c'
in.
Ac
in.2
J/c =
J/c'
in.3
c = c'
in.
10x10 206.3 963 6.88 242.3 1176 7.13 280.3 1414 7.38 320.3 1676 7.63
12x12 236.3 1258 7.88 276.3 1522 8.13 318.3 1813 8.38 362.3 2131 8.63
14x14 266.3 1593 8.88 310.3 1913 9.13 356.3 2262 9.38 404.3 2642 9.63
16x16 296.3 1968 9.88 344.3 2349 10.13 394.3 2763 10.38 446.3 3209 10.63
18x18 326.3 2383 10.88 378.3 2831 11.13 432.3 3314 11.38 488.3 3832 11.63
20x20 356.3 2838 11.88 412.3 3358 12.13 470.3 3915 12.38 530.3 4511 12.63
22x22 386.3 3333 12.88 446.3 3930 13.13 508.3 4568 13.38 572.3 5246 13.63
24x24 416.3 3868 13.88 480.3 4548 14.13 546.3 5271 14.38 614.3 6037 14.63
h = 7 in.,
d = 53/4 in.
h = 71/2 in.,
d = 61/4 in.
h = 8 in.,
d = 63/4 in.
h = 81/2 in.,
d = 71/4 in.
COL.
SIZE
Ac
in.2
J/c =
J/c'
in.3
c = c'
in.
Ac
in.2
J/c =
J/c'
in.3
c = c'
in.
Ac
in.2
J/c =
J/c'
in.3
c = c'
in.
Ac
in.2
J/c =
J/c'
in.3
c = c'
in.
10x10 362.3 1965 7.88 406.3 2282 8.13 452.3 2628 8.38 500.3 3003 8.63
12x12 408.3 2479 8.88 456.3 2857 9.13 506.3 3267 9.38 558.3 3709 9.63
14x14 454.3 3054 9.88 506.3 3499 10.13 560.3 3978 10.38 616.3 4492 10.63
16x16 500.3 3690 10.88 556.3 4207 11.13 614.3 4761 11.38 674.3 5352 11.63
18x18 546.3 4388 11.88 606.3 4982 12.13 668.3 5616 12.38 732.3 6290 12.63
20x20 592.3 5147 12.88 656.3 5824 13.13 722.3 6543 13.38 790.3 7305 13.63
22x22 638.3 5967 13.88 706.3 6732 14.13 776.3 7542 14.38 848.3 8397 14.63
24x24 684.3 6849 14.88 756.3 7707 15.13 830.3 8613 15.38 906.3 9567 15.63
Table 4-7 Properties of Critical Transfer Section-Interior Column
c
c'
c1
c2
b2 = c2+2d/2
b1
=
c1
+
2
d / 2
γvM
Concrete area of critical section:
Ac
= 2(b1
+ b2
)d
Modulus of critical section:
J
c = J
c' = [b1
d(b1
+ 3b2
) + d3] / 3
where:
c = c' = b1
/2

4-40
h = 9 in.,
d = 73/4 in.
h = 91/2 in.,
d = 81/4 in.
h = 10 in.,
d = 83/4 in.
COL.
SIZE
Ac
in.2
J/c =
J/c'
in.3
c = c'
in.
Ac
in.2
J/c =
J/c'
in.3
c = c'
in.
Ac
in.2
J/c =
J/c'
in.3
c = c'
in.
10x10 550.3 3411 8.88 602.3 3851 9.13 656.3 4325 9.38
12x12 612.3 4186 9.88 668.3 4698 10.13 726.3 5247 10.38
14x14 674.3 5043 10.88 734.3 5633 11.13 796.3 6262 11.38
16x16 736.3 5984 11.88 800.3 6656 12.13 866.3 7370 12.38
18x18 798.3 7007 12.88 866.3 7767 13.13 936.3 8572 13.38
20x20 860.3 8112 13.88 932.3 8966 14.13 1006.3 9867 14.38
22x22 922.3 9301 14.88 998.3 10253 15.13 1076.3 11255 15.38
24x24 984.3 10572 15.88 1064.3 11628 16.13 1146.3 12737 16.38
Table 4-8 Properties of Critical Transfer Section—Edge Column—Bending Parallel to Edge
Table 4-7 continued
c
c'
c1
c2
b2 = c2+d/2
b1
=
c1
+
2
d / 2
γvM
Ac
= (b1
+ 2b2
)d
J
c = J
c' = [b1
d(b1
+ 6b2
) + d3] / 6
where:
c = c' = b1
/2
h = 5 in.,
d = 33/4 in.
h = 51/2 in.,
d = 41/4 in.
h = 6 in.,
d = 43/4 in.
h = 61/2 in.,
d = 51/4 in.
COL.
SIZE
Ac
in.2
J/c =
J/c'
in.3
c = c'
in.
Ac
in.2
J/c =
J/c'
in.3
c = c'
in.
Ac
in.2
J/c =
J/c'
in.3
c = c'
in.
Ac
in.2
J/c =
J/c'
in.3
c = c'
in.
10x10 140.6 739 6.88 163.6 891 7.13 187.6 1057 7.38 212.6 1238 7.63
12x12 163.1 983 7.88 189.1 1175 8.13 216.1 1384 8.38 244.1 1609 8.63
14x14 185.6 1262 8.88 214.6 1499 9.13 244.6 1755 9.38 275.6 2029 9.63
16x16 208.1 1576 9.88 240.1 1863 10.13 273.1 2170 10.38 307.1 2497 10.63
18x18 230.6 1926 10.88 265.6 2267 11.13 301.6 2629 11.38 338.6 3015 11.63
20x20 253.1 2310 11.88 291.1 2710 12.13 330.1 3133 12.38 370.1 3581 12.63
22x22 275.6 2729 12.88 316.6 3192 13.13 358.6 3681 13.38 401.6 4197 13.63
24x24 298.1 3183 13.88 342.1 3715 14.13 387.1 4274 14.38 433.1 4861 14.63

4-41
Table 4-8 continued
h = 7 in.,
d = 53/4 in.
h = 71/2 in.,
d = 61/4 in.
h = 8 in.,
d = 63/4 in.
h = 81/2 in.,
d = 71/4 in.
COL.
SIZE
Ac
in.2
J/c =
J/c'
in.3
c = c'
in.
Ac
in.2
J/c =
J/c'
in.3
c = c'
in.
Ac
in.2
J/c =
J/c'
in.3
c = c'
in.
Ac
in.2
J/c =
J/c'
in.3
c = c'
in.
10x10 238.6 1435 7.88 265.6 1649 8.13 293.6 1879 8.38 322.6 2127 8.63
12x12 273.1 1852 8.88 303.1 2113 9.13 334.1 2393 9.38 366.1 2692 9.63
14x14 307.6 2322 9.88 340.6 2635 10.13 374.6 2969 10.38 409.6 3325 10.63
16x16 342.1 2846 10.88 378.1 3216 11.13 415.1 3609 11.38 453.1 4025 11.63
18x18 376.6 3423 11.88 415.6 3855 12.13 455.6 4311 12.38 496.6 4793 12.63
20x20 411.1 4054 12.88 453.1 4552 13.13 496.1 5077 13.38 540.1 5628 13.63
22x22 445.6 4739 13.88 490.6 5308 14.13 536.6 5905 14.38 583.6 6531 14.63
24x24 480.1 5477 14.88 528.1 6122 15.13 577.1 6797 15.38 627.1 7502 15.63
h = 9 in.,
d = 73/4 in.
h = 91/2 in.,
d = 81/4 in.
h = 10 in.,
d = 83/4 in.
COL.
SIZE
Ac
in.2
J/c =
J/c'
in.3
c = c'
in.
Ac
in.2
J/c =
J/c'
in.3
c = c'
in.
Ac
in.2
J/c =
J/c'
in.3
c = c'
in.
10x10 352.6 2393 8.88 383.6 2678 9.13 415.6 2983 9.38
12x12 399.1 3011 9.88 433.1 3351 10.13 468.1 3713 10.38
14x14 445.6 3702 10.88 482.6 4101 11.13 520.6 4524 11.38
16x16 492.1 4464 11.88 532.1 4928 12.13 573.1 5417 12.38
18x18 538.6 5299 12.88 581.6 5832 13.13 625.6 6392 13.38
20x20 585.1 6207 13.88 613.1 6814 14.13 678.1 7449 14.38
22x22 631.6 7187 14.88 680.6 7872 15.13 730.6 8587 15.38
24x24 678.1 8239 15.88 730.1 9007 16.13 783.1 9807 16.38

4-42
Table 4-9 Properties of Critical Transfer Section—Edge Column—Bending Perpendicular to Edge
c
c'
c1 c2
b2 = c2+2d/2
b1
=
c1
+
d / 2
γvM
Ac
= (2b1
+ b2
)d
J
c = [2b1
d(b1
+ 2b2
) + d3(2b1
+ b2
)/ b1
]/6
J
c' = [2b1
2d(b1
+ 2b2
) + d3(2b1
+ b2
)]/6(b1
+ b2
)
where:
c = b1
2/(2b1
+ b2
)
c' = b1
(b1
+ b2
)/(2b1
+ b2
)
h = 5 in.,
d = 33/4 in.
h = 51/2 in.,
d = 41/4 in.
h = 6 in.,
d = 43/4 in.
COL.
SIZE
Ac
in.2
J/c
in.3
J/c'
in.3
c
in.
c'
in.
Ac
in.2
J/c
in.3
J/c'
in.3
c
in.
c'
in.
Ac
in.2
J/c
in.3
J/c'
in.
3
c
in.
c'
in.
10x10 140.6 612 284 3.76 8.11 163.6 738 339 3.82 8.31 187.6 878 400 3.88 8.50
12x12 163.1 815 381 4.43 9.45 189.1 973 453 4.48 9.64 216.1 1146 529 4.54 9.83
14x14 185.6 1047 494 5.09 10.78 214.6 1242 583 5.15 10.98 244.6 1453 677 5.21 11.17
16x16 208.1 1309 622 5.76 12.12 240.1 1545 730 5.81 12.31 273.1 1798 844 5.87 12.50
18x18 230.6 1602 765 6.42 13.45 265.6 1882 894 6.48 13.64 301.6 2181 1030 6.54 13.84
20x20 253.1 1924 923 7.09 14.79 291.1 2253 1075 7.15 14.98 330.1 2602 1235 7.20 15.17
22x22 275.6 2277 1095 7.76 16.12 316.6 2658 1273 7.81 16.31 358.6 3061 1459 7.87 16.51
24x24 298.1 2659 1283 8.42 17.45 342.1 3097 1488 8.48 17.65 387.1 3558 1702 8.54 17.84
h = 6 1/2 in.,
d = 51/4 in.
h = 7 in.,
d = 53/4 in.
h = 71/2 in.,
d = 61/4 in.
COL.
SIZE
Ac
in.2
J/c
in.3
J/c'
in.3
c
in.
c'
in.
Ac
in.2
J/c
in.3
J/c'
in.3
c
in.
c'
in.
Ac
in.2
J/c
in.3
J/c'
in.3
c
in.
c'
in.
10x10 212.6 1030 467 3.94 8.69 238.6 1197 538 3.99 8.88 265.6 1379 616 4.05 9.07
12x12 244.1 1334 612 4.60 10.03 273.1 1537 701 4.66 10.22 303.1 1757 796 4.72 10.41
14x14 275.6 1680 779 5.26 11.36 307.6 1924 886 5.32 11.55 340.6 2185 1001 5.38 11.74
16x16 307.1 2068 966 5.93 12.70 342.1 23.56 1095 5.99 12.89 378.1 2664 1231 6.05 13.08
18x18 338.6 2498 1174 6.60 14.03 376.6 2835 1326 6.65 14.22 415.6 3192 1486 6.71 14.41
20x20 370.1 2970 1404 7.26 15.36 411.1 3360 1581 7.32 15.56 453.1 3771 1766 7.38 15.75
22x22 401.6 3485 1654 7.93 16.70 445.6 3931 1858 7.98 16.89 490.6 4400 2071 8.04 17.08
24x24 433.1 4041 1926 8.59 18.03 480.1 4548 2158 8.65 18.23 528.1 5078 2401 8.71 18.42

4-43
h = 8 in.,
d = 63/4 in.
h = 81/2 in.,
d = 71/4 in.
h = 9 in.,
d = 73/4 in.
COL.
SIZE
Ac
in.2
J/c
in.3
J/c'
in.3
c
in.
c'
in.
Ac
in.2
J/c
in.3
J/c'
in.3
c
in.
c'
in.
Ac
in.2
J/c
in.3
J/c'
in.3
c
in.
c'
in.
10x10 293.6 1577 700 4.11 9.26 322.6 1792 791 4.17 9.45 352.6 2024 888 4.23 9.64
12x12 334.1 1994 898 4.78 10.6 366.1 2249 1008 4.83 10.79 399.1 2523 1124 4.89 10.98
14x14 374.6 2465 1124 5.44 11.94 409.6 2765 1253 5.50 12.13 445.6 3084 1391 5.56 12.32
16x16 415.1 2991 1376 6.10 13.27 453.1 3338 1528 6.16 13.46 492.1 3707 1689 6.22 13.65
18x18 455.6 3571 1655 6.77 14.61 496.6 3970 1832 6.83 14.80 538.6 4393 2018 6.89 14.99
20x20 496.1 4204 1961 7.43 15.94 540.1 4661 2164 7.49 16.13 585.1 5141 2378 7.55 16.33
22x22 536.6 4892 2294 8.1 17.28 583.6 5409 2526 8.16 17.47 631.6 5951 2768 8.21 17.66
24x24 577.1 5634 2654 8.76 18.61 627.1 6216 2916 8.82 18.80 678.1 6823 3190 8.88 18.99
h = 91/2 in.,
d = 81/4 in.
h = 10 in.,
d = 83/4 in.
COL.
SIZE
Ac
in.2
J/c
in.3
c
in.
c'
in.
Ac
in.2
J/c
in.3
J/c'
in.3
c
in.
c'
in.
Ac
in.2
10x10 383.6 2275 992 4.29 9.83 415.6 2544 1104 4.35 10.02
12x12 433.1 2816 1248 4.95 11.17 468.1 3129 1380 5.01 11.36
14x14 482.6 3424 1537 5.62 12.51 520.6 3785 1691 5.67 12.70
16x16 532.1 4098 1858 6.28 13.85 573.1 4511 2037 6.34 14.04
18x18 581.6 4839 2213 6.94 15.18 625.6 5308 2418 7.00 15.37
20x20 631.1 5646 2601 7.61 16.52 678.1 6176 2834 7.67 16.71
22x22 680.6 6519 3021 8.27 17.85 730.6 7113 3284 8.33 18.04
24x24 730.1 7458 3474 8.94 19.19 783.1 8121 3770 9.00 19.38
Table 4-9 continued

4-44
Table 4-10 Properties of Critical Transfer Section—Corner Column
c
c'
c1
c2
b2 = c2+d/2
b1
=
c1
+
d / 2
γvM
Ac
= (b1
+ b2
)d
J
c = [2b1
d(b1
+ 2b2
) + d3(2b1
+ b2
)/ b1
]/6
J
c' = [2b1
2d(b1
+ 2b2
) + d3(2b1
+ b2
)]/6(b1
+ b2
)
where:
c = b1
2/(2b1
+ b2
)
c' = b1
(b1
+ b2
)/(2b1
+ b2
)
h = 5 in.,
d = 33/4 in.
h = 51/2 in.,
d = 41/4 in.
h = 6 in.,
d = 43/4 in.
COL.
SIZE
Ac
in.2
J/c
in.3
J/c'
in.3
c
in.
c'
in.
Ac
in.2
J/c
in.3
J/c'
in.3
c
in.
c'
in.
Ac
in.2
J/c
in.3
J/c'
in.3
c
in.
c'
in.
10x10 89.1 458 153 2.97 8.91 103.1 546 182 3.03 9.09 117.6 642 214 3.09 9.28
12x12 104.1 619 206 3.47 10.41 120.1 732 244 3.53 10.59 136.6 854 285 3.59 10.78
14x14 119.1 805 268 3.97 11.91 137.1 946 315 4.03 12.09 155.6 1097 366 4.09 12.28
16x16 134.1 1016 339 4.47 13.41 154.1 1189 396 4.53 13.59 174.6 1372 457 4.59 13.78
18x18 149.1 1252 417 4.97 14.91 171.1 1460 487 5.03 15.09 193.6 1679 560 5.09 15.28
20x20 164.1 1513 504 5.47 16.41 188.1 1759 586 5.53 16.59 212.6 2017 672 5.59 16.78
22x22 179.1 1799 600 5.97 17.91 205.1 2087 696 6.03 18.09 231.6 2388 796 6.09 18.28
24x24 194.1 2110 703 6.47 19.41 222.1 2443 814 6.53 19.59 250.6 2789 930 6.59 19.78
h = 61/2 in.,
d = 51/4 in.
h = 7 in.,
d = 53/4 in.
h = 71/2 in.,
d = 61/4 in.
COL.
SIZE
Ac
in.2
J/c
in.3
J/c'
in.3
c
in.
c'
in.
Ac
in.2
J/c
in.3
J/c'
in.3
c
in.
c'
in.
Ac
in.2
J/c
in.3
J/c'
in.3
c
in.
c'
in.
10x10 132.6 746 249 3.16 9.47 148.1 858 286 3.22 9.66 164.1 979 326 3.28 9.84
12x12 153.6 984 328 3.66 10.97 171.1 1124 375 3.72 11.16 189.1 1273 424 3.78 11.34
14x14 174.6 1257 419 4.16 12.47 194.1 1428 476 4.22 12.66 124.1 1609 536 4.28 12.84
16x16 195.6 1566 522 4.66 13.97 217.1 1770 590 4.72 14.16 239.1 1986 662 4.78 14.34
18x18 216.6 1909 636 5.16 15.47 240.1 2151 717 5.22 15.66 264.1 2406 802 5.28 15.84
20x20 237.6 2288 763 5.66 16.97 263.1 2571 857 5.72 17.16 289.1 2867 956 5.78 17.34
22x22 258.6 2701 900 6.16 18.47 286.1 3028 1009 6.22 18.66 314.1 3369 1123 6.28 18.84
24x24 279.6 3150 1050 6.66 19.97 309.1 3524 1175 6.72 20.16 339.1 3913 1304 6.78 20.34

4-45
h = 8 in.,
d = 63/4 in.
h = 81/2 in.,
d = 71/4 in.
h = 9 in.,
d = 73/4 in.
COL.
SIZE
Ac
in.2
J/c
in.3
J/c'
in.3
c
in.
c'
in.
Ac
in.2
J/c
in.3
J/c'
in.3
c
in.
c'
in.
Ac
in.2
J/c
in.3
J/c'
in.3
c
in.
c'
in.
10x10 180.6 1109 370 3.34 10.03 197.6 1249 416 3.41 10.22 215.1 1398 466 3.47 10.41
12x12 207.6 1432 477 3.84 11.53 226.6 1602 534 3.91 11.72 246.1 1783 594 3.97 11.91
14x14 234.6 1801 600 4.34 13.03 255.6 2004 668 4.41 13.22 277.1 2219 740 4.47 13.41
16x16 261.6 2214 738 4.84 14.53 284.6 2454 818 4.91 14.72 308.1 2706 902 4.97 14.91
18x18 288.6 2673 891 5.34 16.03 313.6 2952 984 5.41 16.22 339.1 3246 1082 5.47 16.41
20x20 315.6 3176 1059 5.84 17.53 342.6 3499 1166 5.91 17.72 370.1 3837 1279 5.97 17.91
22x22 342.6 3724 1241 6.34 19.03 371.6 4094 1365 6.41 19.22 401.1 4479 1493 6.47 19.41
24x24 369.6 4318 1439 6.84 20.53 400.6 4738 1579 6.91 20.72 432.1 5173 1724 6.97 20.91
h = 91/2 in.,
d = 81/4 in.
h = 10 in.,
d = 83/4 in.
COL.
SIZE
Ac
in.2
J/c
in.3
J/c'
in.3
c'
in.
Ac
in.2
J/c
in.3
J/c'
in.3
c
in.
c'
in.
Ac
in.2
10x10 233.1 1559 520 3.53 10.59 251.6 1730 577 3.59 10.78
12x12 266.1 1975 658 4.03 12.09 286.6 2178 726 4.09 12.28
14x14 299.1 2446 815 4.53 13.59 321.6 2685 895 4.59 13.78
16x16 332.1 2972 991 5.03 15.09 356.6 3250 1083 5.09 15.28
18x18 365.1 3553 1184 5.53 16.59 391.6 3874 1291 5.59 16.78
20x20 398.1 4189 1396 6.03 18.09 426.6 4556 1519 6.09 18.28
22x22 431.1 4879 1626 6.53 19.59 461.6 5296 1765 6.59 19.78
24x24 464.1 5625 1875 7.03 21.09 496.6 6094 2031 7.09 21.28
Table 4-10 continued
References
4.1 Notes on ACI 318-08, Chapter 10: Deflections, EB708, Portland Cement Association, Skokie, Illinois, 2008.
4.2 “Aspects of Design of Reinforced Concrete Flat Plate Slab Systems,” by S.K. Ghosh, Analysis and Design
of High-Rise Concrete Buildings, SP-97, American Concrete Institute, Detroit, Michigan, 1985, pp. 139-157.

4-46

5-1
Chapter 5
Simplified Design for Columns
5.1 INTRODUCTION
Use of high-strength materials has had a significant effect on the design of concrete columns. Increased use of
high-strength concretes has resulted in columns that are smaller in size and, therefore, are more slender.
Consequently, in certain situations, slenderness effects must be considered, resulting in designs that are more
complicated than when these effects may be neglected.
For buildings with adequate shearwalls, columns may be designed for gravity loads only. However, in some
structures—especially low-rise buildings—it may not be desirable or economical to include shearwalls.
In these situations, the columns must be designed to resist both gravity and lateral loads. In either case, it is
important to be able to distinguish between a column that is slender and one that is not. A simplified design
procedure is outlined in this chapter, which should be applicable to most columns. Design aids are given to
assist the engineer in designing columns within the limitations stated.
5.2 DESIGN CONSIDERATIONS
5.2.1 Column Size
The total loads on columns are directly proportional to the bay sizes (i.e. tributary areas). Larger bay sizes mean
more load to each column and, thus, larger column sizes. Bay size is often dictated by the architectural and
functional requirements of the building. Large bay sizes may be required to achieve maximum unobstructed
floor space. The floor system used may also dictate the column spacing. For example, the economical use of a
flat plate floor system usually requires columns that are spaced closer than those supporting a pan joist floor
system. Architecturally, larger column sizes can give the impression of solidity and strength, whereas smaller
columns can express slender grace. Aside from architectural considerations, it is important that the columns
satisfy all applicable strength requirements of the ACI Code, and at the same time, be economical. Minimum
column size and concrete cover to reinforcement may be governed by fire-resistance criteria (see Chapter 10,
Tables 10-2 and 10-6).
5.2.2 Column Constructability
Columns must be sized not only for adequate strength, but also for constructability. For proper concrete
placement and consolidation, the engineer must select column size and reinforcement to ensure that the
reinforcement is not congested. Bar lap splices and locations of bars in beams and slabs framing into the column
must be considered. Columns designed with a smaller number of larger bars usually improve constructability.

5.2.3 Column Economics
Concrete is more cost effective than reinforcement for carrying compressive axial load; thus, it is more
economical to use larger column sizes with lesser amounts of reinforcement. Also, columns with a smaller
number of larger bars are usually more economical than columns with a larger number of smaller bars.
Reuse of column forms from story level to story level results in significant savings. It is economically unsound
to vary column size to suit the load on each story level. It is much more economical to use the same column
size for the entire building height, and to vary only the longitudinal reinforcement as required. In taller
buildings, the concrete strength is usually varied along the building height as well.
5.3 DESIGN STRENGTH FOR COLUMNS
For columns subjected to axial loads only (with no or negligible bending moment) the code provides the
following equations for the design axial strength φPn
:
For columns with spiral reinforcement conforming to Section 10.9.3
φPn, max
= 0.85φ[0.85 (Ag
-Ast
) + fy
Ast
] ACI Eq. (10-1)
For columns with tie reinforcement conforming to Section 7.10.5
φPn, max
= 0.80φ[0.85 (Ag
-Ast
) + fy
Ast
] ACI Eq. (10-2)
Were Ag
and Ast
are the gross area of the column cross section and reinforcement area respectively.
The strength reduction factor φ is taken as 0.75 for columns with spiral reinforcement and 0.65 for with
tie reinforcement.
In general columns are subjected to combined axial load and bending moment, Pu
and Mu
. The design strength
for a column cross section in this case is expressed by interaction diagram representing all possible
combinations of the design strengths φPn
and φMn
for a specific cross section. To develop an interaction
diagram two conditions must be satisfied: static equilibrium and compatibility of strain. Different strain
profiles are assumed, each strain profile is associated with one point on the interaction diagram. For each strain
profile the internal forces acting on the cross section are calculated using the assumptions introduced in Chapter
3 (See Figure 3-2). Figure 5-1, illustrates the development of interaction diagram for nominal strength Pn
and
Mn
. To develop the interaction diagram for design strength (φPn
and φMn
), the nominal strength is multiplied
by the strength reduction factor φ. The value of φ depends on whether the column cross section is tension
controlled, compression controlled or in transition between these two limits. The definition of tension and
compression controlled section depend on the magnitude of the net tensile strain at the extreme layer of
longitudinal tension steel at nominal strength. The strength reduction φ is calculated as follows:
1. For compression controlled section (εt
≤ 0.002) φ = 0.65 for columns with tie reinforcement and
φ = 0.75 for columns with spiral reinforcement
2. For tension controlled section (εt
≥ 0.005) φ = 0.9
ʹfc
ʹfc
5-2

5-3
Chapter 5 • Simplified Design for Columns

!
# $ !
% !' () * +'* , !

(! ( '+'

-
- (*'* ' ) ' ./ 0

(! ' ( '+'

*( ') '
1
-
* ( ') ' (
Pn
Mn

Pn

Mn

Figure 5-1 Development of Interaction Diagram for Nominal Strength (Pn and Mn)

3. For other values of εt
(transition, 0.005 εt
0.002 ) φ is calculated as follows:
φ = 0.65 + (εt
– 0.002)(250/3) for columns with tie reinforcement
φ = 0.75 + (εt
– 0.002)(250) for columns with spiral reinforcement
It is important to note that the value for φPn
should not exceed the value of φPn
, max
as calculate from ACI
equation (10-1) or (10-2) above.
For columns subjected to combined axial load and biaxial bending (Pu
, Mux
and Muy
), the design strength is
expressed by three-dimensional interaction surface. A simplified design method for this case is introduced in
Section 5.5.3.
5-4
5.4 PRELIMINARY COLUMN SIZING
It is necessary to select a preliminary column size for cost estimating and/or frame analysis. The initial selection
can be very important when considering overall design time. In general, a preliminary column size should be
determined using a low percentage of reinforcement; it is then possible to provide any additional reinforcement
required for the final design (including applicable slenderness effects) without having to change the column size.
Columns which have reinforcement ratios in the range of 1% to 2% will usually be the most economical.
The design charts presented in Figures 5-2 and 5-3, are based on ACI Eq. (10-2). These charts can be used for
nonslender tied square columns loaded at an eccentricity of no more than 0.1h, where h is the size of the
column. Design axial load strengths φPn(max)
for column sizes from 10 in. to 24 in. with reinforcement ratios
between 1 and 8% are presented in Fig. 5-2. For other columns sizes and shapes, and concrete strengths the
chart in Fig. 5-3, based on ACI Eq (10-2) can be used for preliminary column sizing.
These design charts will provide quick estimates for a column size required to support a factored load Pu
within
the allowable limits of the reinforcement ratio (ACI 10.9). Using the total tributary factored load Pu
for
the lowest story of a multistory column stack, a column size should be selected with a low percentage of
reinforcement. This will allow some leeway to increase the amount of steel for the final design, if required.

5-5
0
200
400
600
800
1000
1200
1400
1600
1800
2000
2200
2400
1 2 3 4 5 6 7 8
ρg = Ast /Ag, %
φPn(max)=Pu,kips
24 x 24
22 x 22
20 x 20
20 x 20
18 x 18
16 x 16
14 x 14
12 x 12
10 x 10
φPn(max) = 0.80φAg 0.85fc' + ρg fy − 0.85fc'( )[ ]
φ = 0.65
fc' = 4000 psi
fy = 60,000 psi
Figure 5-2 Design Chart for Nonslender, Square Tied Columns

5-6
7
6.5
6
5.5
5
4.5
4
3.5
3
2.5
2
1 2 3 4 5 6
ρ = Ast/Ag
Pu/Ag
ʹfc
= 12 ksi
11 ksi
10 ksi
9 ksi
8 ksi
7 ksi
6 ksi
5 ksi
4 ksi
Figure 5-3 Design Chart for Nonslender, Tied Columns

5.5 SIMPLIFIED DESIGN FOR COLUMNS
5.5.1 Simplified Design Charts—Combined Axial Load and Bending Moment
Numerous design aids and computer programs are available for determining the size and reinforcement of columns
subjected to axial forces and/or moments. Tables, charts, and graphs provide design data for a wide variety of
column sizes and shapes, reinforcement layouts, load eccentricities and other variables. These design aids
eliminate the necessity for making complex and repetitious calculations to determine the strengths of columns, as
preliminarily sized. The design aids presented in References 5.1 and 5.2 are widely used. In addition, extensive
column load tables are available in the CRSI Handbook. Each publication presents the design data in a somewhat
different format; however, the accompanying text in each reference readily explains the method of use. Computer
programs may also be used to design or investigate rectangular or circular column sections with any
reinforcement layout or pattern.
In general, columns must be designed for the combined effects of axial load and bending moment. As noted
earlier, appreciable bending moments can occur in columns because of unbalanced gravity loads and/or lateral
forces. Simplified interaction diagram, such as the one depicted in Figure 5-4, can be created. For simplicity
the diagram is plotted by connecting straight lines between points corresponding to certain transition stages.
The transition stages are as follows (see Fig. 5-4):
Stage 1: Pure compression (no bending moment)
Stage 2: Stress in reinforcement closest to tension face = 0 (fs = 0)
Stage 3: Stress in reinforcement closest to tension face = 0.5 fy (fs = 0.5 fy)
Stage 4: Balanced point; stress in reinforcement closest to tension face = fy (fs = fy)
Stage 5: Pure bending (no axial load)
Note that Stages 2 and 3 are used to determine which type of lap splice is required for a given load combina-
tion (ACI 12.17). In particular, for load combinations falling within Zone 1, compression lap splices are
allowed, since all of the bars are in compression. In Zone 2, either Class A (half or fewer of the bars spliced at
one location) or Class B (more than one-half of the bars spliced at one location) tension lap splices must be
used. Class B tension lap splices are required for load combinations falling within Zone 3.
For the general case of a rectangular column section (Fig. 5-4):
1. Point 1
φPn(max)
= 0.80Ag
[0.85 + ρg
(fy
− 0.85 )] (kips)
Where Ag
= gross area of column, in2
Ast
= total area of longitudinal reinforcement, in2
ρg
= Ast
/Ag
φ = strength reduction factor = 0.65
ʹfc ʹfc
5-7

2. Points 2-4
(kips)
(ft-kips)
where
To ensure that the stress in the reinforcement at each layer ≤ fy
= 60 ksi:
Where :
h = column dimension in the direction of bending, in.
b = column dimension perpendicular to the direction of bending
d1
= distance from compression face to centroid of reinforcing steel in layer 1 (layer closest to tension side) in.
di
= distance from compression face to centroid of reinforcing steel in layer i in.
Asi
= total steel area in layer i, in.2
n = total number of layers of reinforcement
εs1
= steel strain in layer 1
εi
= steel strain in layer i
Values for C1
and C2
are presented in Table 5-1
Point 5 Pure bending
In lieu of Iterative procedure to determine φMn, the simplified approach introduced in Chapter 3
may be used.
Table 5-1 Constants for strain compatibility analysis – Rectangular section
(ksi) 4 5 6 7 8 9 10 11 12 C2 for fs1
/fy
εs1
all
Point 2 2.89 3.40 3.83 4.17 4.42 4.97 5.53 6.08 6.63 1.00 0 0
C1 Point 3 2.15 2.53 2.84 3.10 3.29 3.70 4.11 4.52 4.93 1.34 -0.5 -0.001
Point 4 1.71 2.01 2.26 2.47 2.62 2.94 3.27 3.60 3.92 1.69 -1 -0.002
β1
0.85 0.8 0.75 0.7 0.65 0.65 0.65 0.65 0.65
ʹfc
ʹfc
0.85 ≥ β1
= 1.05 − 0.05 ʹfc
≥ 0.65
1− C2
di
d1
⎛
⎝
⎜
⎞
⎠
⎟ ≤
60
87
= 0.69
C2 =
0.003− εS1
0.003
C1
= 0.85 ʹfc
β1
0.003
0.003− εS1
⎛
⎝
⎜
⎞
⎠
⎟
φMn
= φ 0.5C1
d1
b h −
β1
d1
C2
⎛
⎝
⎜
⎞
⎠
⎟ + 87∑n
i =1Asi
1− C2
di
d1
⎛
⎝
⎜
⎞
⎠
⎟
h
2
− di
⎛
⎝
⎜
⎞
⎠
⎟
⎡
⎣
⎢
⎤
⎦
⎥ /12
Pn
= φ C1
d1
b + 87∑n
i =1Asi
1− C2
di
d1
⎛
⎝
⎜
⎞
⎠
⎟
⎡
⎣
⎢
⎤
⎦
⎥φ
5-8

For columns with square sections and concrete strength = 4 ksi, the above equations further simplify
as follows;
(1) Point 1 (see Fig. 5-2):
φPn(max) = 0.80φ[0.85 (Ag – Ast) + fyAst] ACI Eq. (10-2)
= 0.80φAg
[0.85 + ρg
(fy
– 0.85 )]
(2) Points 2-4 (see Fig. 5-4):
(5-1)
(5-2)φMn
= φ 0.5C1
hd1
h − C3
d1( )+ 87 Asi
1− C2
di
d1
⎛
⎝
⎜
⎞
⎠
⎟
h
2
⎛
⎝
⎜
⎞
⎠
⎟ − di
i=1
n
∑
⎡
⎣
⎢
⎤
⎦
⎥ /12
φPn
= φ C1
hd1
+ 87 Asi
1− C2
di
d1
⎛
⎝
⎜
⎞
⎠
⎟
i=1
n
∑
⎡
⎣
⎢
⎤
⎦
⎥
ʹfcʹfc
ʹfc
ʹfc
5-9
Axial Load
All bars in compression
Zone 1
Zone 2
Zone 3
0 ≤ fs ≤ 0.5fy
on tension face
fs ≥ 0.5fy
on tension face
1
2
3
4
5
Bending Moment
*
Figure 5-4 Transition Stages on Interaction Diagram
Layer n, Asn
Layer i, Asi
Layer 1, As1
Compression face
Tension face
b
dn
di
d1 h
0.003
c
εsn
εsi
εs1
Strain diagram
*0.65[0.85 (Ag
– Ast
) + fy
Ast
]ʹfc

(3) Point 5:
For columns with 2 or 3 layers of reinforcement:
φMn
= 4As1
d1
For columns with 4 or 5 layers of reinforcement:
φMn
= 4(As1
+ As2
)(d1
–
s
2 )
where s = center to center spacing of the bars
In both equations, φ = 0.90; also, As1 and As2 are in.2
, d1 and s are in in., and φMn is in ft-kips.
The simplified equations for Points 2-4 will produce values of φPn
and φMn
approximately 3% larger than the
exact values (at most). The equations for Point 5 will produce conservative values of φMn for the majority of
cases. For columns subjected to small axial loads and large bending moments, a more precise investigation into
the adequacy of the section should be made because of the approximate shape of the simplified interaction
diagram in the tension-controlled region. However, for typical building columns, load combinations of this type
are rarely encountered.
5-10
h
Compression Face
Layer n, Asn
Layer i, Asi
Layer 1, As1
dn
h
di
d1
1.5 (typ.)Tension Face
Figure 5-5 Notation for Eqs. (5-1) and (5-2)
Table 5-2 Constants for Points 2-4
Point No. C1 C2 C3
2 2.89 1.00 0.85
3 2.14 1.35 0.63
4 1.70 1.69 0.50

To allow rapid selection of column size and longitudinal reinforcement for a factored axial load Pu
and bending
moment Mu
, Figs. 5-18 through 5-25 are included at the end of this chapter. All design charts are based on
= 4000 psi and fy
= 60,000 psi, and are valid for square, tied, nonslender columns with symmetrical bar
arrangements as shown in Fig. 5-6. The number in parentheses next to the number of reinforcing bars is the
reinforcement ratio, ρg
= Ast
/Ag
, where Ast
is the total area of the longitudinal bars and Ag
is the gross area of
the column section. A clear cover of 1.5 in. to the ties was used (ACI 7.7.1); also used was No. 3 ties with
longitudinal bars No. 10 and smaller and No. 4 ties with No. 11 bars (ACI 7.10.5).
For a column with a larger cross-section than required for loads, a reduced effective area not less than one-half
of the total area may be used to determine the minimum reinforcement and the design capacity (ACI 10.8.4),
this provision must not be used in regions of high seismic risk.
Essentially this means that a column of sufficient size can be designed to carry the design loads, and concrete
added around the designed section without having to increase the amount of longitudinal reinforcement to
satisfy the minimum requirement in ACI 10.9.1. Thus, in these situations, the minimum steel percentage, based
on actual gross cross-sectional area of column, may be taken less than 0.01, with a lower limit of 0.005 (the
exact percentage will depend on the factored loads and the dimensions of the column). It is important to note
that the additional concrete must not be considered as carrying any portion of the load, but must be considered
when computing member stiffness (ACI R10.8.4).
Additional design charts for other column sizes and material strengths can obviously be developed. For
rectangular or round columns, the graphs presented in Reference 5.2 may be used; these graphs are presented
in a nondimensionalized format and cover an extensive range of column shapes and material strengths. Also,
the CRSI Handbook5.3
gives extensive design data for square, rectangular, and round column sections.
5.5.1.1 Example: Construction of Simplified Design Chart
To illustrate the simplified procedure for constructing column interaction diagrams, determine the points
corresponding to the various transition stages for an 18 × 18 in. column reinforced with 8-No. 9 bars, as shown
in Fig. 5-7.
ʹfc
5-11
4 bars
8 bars
12 bars
16 bars
Figure 5-6 Bar Arrangements for Column Design Charts

(1) Point 1 (pure compression):
= 0.0247
φPn(max)
= (0.80 × 0.65 × 324)[(0.85 × 4) + 0.0247(60-(0.85 × 4))]
= 808 kips
(2) Point 2 (fs1
= 0):
Using Fig. 5-7 and Table 5-2:
Layer 1:
Layer 2:
Layer 3:
1–C2
d3
/d1
being greater than 0.69 in layer 3 means that the steel in layer 3 has yielded; therefore, use
1–C2d3/d1 = 0.69.
ρg =
8.0
18 ×18
1− C2
d2
d1
= 1−1×
9.00
15.56
⎛
⎝
⎜
⎞
⎠
⎟ = 0.42
1− C2
d1
d1
= 1−1 = 0
1− C2
d3
d1
= 1−1×
2.44
15.56
⎛
⎝
⎜
⎞
⎠
⎟ = 0.84 0.69
5-12
18
#3 tie
1.5
(typ)
d3=2.44
d2=9.00
d1=15.66
18
3-#9
As3 = 3.0 in.2
2-#9
As2 = 2.0 in.2
3-#9
As1 = 3.0 in.2
Figure 5-7 Column Cross-Section for Example Problem

φPn = 0.65
= 0.65{(2.89 × 18 × 15.56) + 87[(3 × 0) +(2 × 0.42) +(3 × 0.69)]}
= 0.65 (809.4 + 253.2)
= 690 kips
φMn = 0.65
= 0.65{(0.5 × 2.89 ×18 ×15.56)[18 – (0.85 ×15.56)]
+ 87[(3.0 × 0(9 – 15.56)) + (2.0 × 0.42(9 – 9)) + (3.0 × 0.69(9 – 2.44))]}/12
= 0.65 (1932.1 +1181.4)/12
= 169 ft-kips
(3) Point 3 (fs1
= 0.5 fy
):
In this case, C1
= 2.14, C2
= 1.35, and C3
= 0.63 (Table 5-1) (Table 5-1)
Layer 1:
Layer 2:
Layer 3:
φPn
= 0.65{(2.14 ×18 ×15.56) + 87[(3 – (-0.35)) + (2 ×0.22) + (3 × 0.69)]}
= 0.65 (599.4 + 127.0) = 472 kips
φMn
= 0.65{(0.5 ×2.14 ×18 ×15.56)[18 – (0.63 ×15.56)]
+ 87[(3.0(-0.35) ×(9 – 15.56)) + 0 + (3.0 ×0.69(9 – 2.44))]}/12
= 0.65 (2456.6 + 1780.7)/12 = 229 ft-kips
(4) Point 4 (fs1
= fy
):
In this case, C1
= 1.70, C2
= 1.69, and C3
= 0.50
Similar calculations yield the following:
φPn
= 312 kips
φMn
= 260 ft-kips
C1
hd1
+ 87 Asi
1− C2
di
d1
⎛
⎝
⎜
⎞
⎠
⎟
i=1
3
∑
⎡
⎣
⎢
⎤
⎦
⎥
0.5C1
hd1
C3
d1( )+ 87 Asi
1− C2
di
d1
⎛
⎝
⎜
⎞
⎠
⎟
h
2
− di
⎛
⎝
⎜
⎞
⎠
⎟
i=1
3
∑
⎡
⎣
⎢
⎤
⎦
⎥
1− C2
d2
d1
= 1−1.35
2.44
15.56
⎛
⎝
⎜
⎞
⎠
⎟ = 0.79 0.69 Use 0.69
1− C2
d2
d1
= 1−1.35
9.00
15.56
⎛
⎝
⎜
⎞
⎠
⎟ = −0.22
1− C2
d1
d1
= 1−1.35 = −0.35
5-13

5-14
Figure 5-8 Comparison of Simplified and spColumn Interation Diagrams
P (kip)
(Pmax)
1000
600
200
-200
-600
1
2
3
4
5
100 200 300
Mx (k-ft)
(Pmin)

5-15
(5) Point 5 (pure bending):
For columns with 3 layers of reinforcement:
φMn
= 4As1
d1
= 4 x 3.0 × 15.56 = 187 ft-kips
Each of these points, connected by straight dotted lines, is shown in Fig. 5-8. The solid line represents the exact
interaction diagram determined from spColumn. As can be seen, the simplified interaction diagram compares
well with the one from spColumn except in the region where the axial load is small and the bending moment
is large; there, the simplified diagram is conservative. However, as noted earlier, typical building columns
will rarely have a load combination in this region. Note that spColumn also gives the portion of the interaction
diagram for tensile axial loads (negative values of φPn
) and bending moments.
Simplified interaction diagrams for all of the other columns in Figs. 5-18 through 5-25 will compare just as well
with the exact interaction diagrams; the largest discrepancies will occur in the region near pure bending only.
5.5.2 Column Ties
The column tie spacing requirements of ACI 7.10.5 are summarized in Table 5-3. For No. 10 column bars and
smaller, No. 3 or larger ties are required; for bars larger than No. 10, No. 4 or larger ties must be used.
Maximum tie spacing shall not exceed the lesser of 1) 16 longitudinal bar diameters, 2) 48 tie bar diameters,
and 3) the least column dimension
Suggested tie details to satisfy ACI 7.10.5.3 are shown in Fig. 5-9 for the 8, 12, and 16 column bar arrangements.
In any square (or rectangular) bar arrangement, the four corner bars are enclosed by a single one-piece tie (ACI
7.10.5.3). The ends of the ties are anchored by a standard 90° or 135° hook (ACI 7.1.3). It is important to alter-
nate the position of hooks in placing successive sets of ties. For easy field erection, the intermediate bars in the
s/2
s/2
tiespacing,s
s/2
3 in.
max.
Interior column-beam joint***
Interior column-slab joint**
*Maximum spacing not to exceed least column dimension (ACI 7.10.5.2)
**Also valid for joints with beams on less than 4 sides of the column (ACI 7.10.5.4)
***Beams on all 4 sides of the column (ACI 7.10.5.5)
Tie
Size
Column
Bars
Maximum
Spacing*
(in.)
#3 # 5 10
# 6 12
# 7 14
# 8 16
# 9 18
#10 18
#4 #11 22
Table 5-3 Column Tie Spacing

5-16
8 and 16 bar arrangements can be supported by the separate crossties shown in Fig. 5-9. Again, it is important
to alternate the position of the 90° hooked end at each successive tie location. The two-piece tie shown for the
12 bar arrangement should be lap spliced at least 1.3 times the tensile development length of the tie bar, ˜d
, but
not less than 12 in. To eliminate the supplementary ties for the 8, 12, and 16 bar arrangements, 2, 3, and 4 bar
bundles at each corner may also be used; at least No. 4 ties are required in these cases (ACI 7.10.5.1).
Column ties must be located not more than one-half a tie spacing above top of footing or slab in any story,
and not more than one-half a tie spacing below the lowest reinforcement in the slab (or drop panel) above (see
ACI 7.10.5.4 and Table 5-2). Where beams frame into a column from four sides, ties may be terminated 3 in.
below the lowest beam reinforcement (ACI 7.10.5.5). Note that extra ties are required within 6 in. from points
of offset bends at column splices (see ACI 7.8.1 and Chapter 8).
Alt. hook
90∞ (typ.)
Column β 18 in. Preassembled
Cages
Field Erection
Alt. hooks (typ.)
20 in., 22 in., and 24 in. columns
8 bars
12 bars
Field Erection All 12 bar arrangements
Lap splice ≥ greater of
16 bars
Preassembled
Cages
Field Erection
All 16 bar arrangements
1.3˜d
12
Figure 5-9 Column Tie Details

5-17
5.5.3 Biaxial Bending of Columns
Biaxial bending of a column occurs when the loading causes bending simultaneously about both principal axes.
This problem is often encountered in the design of corner columns.
A general biaxial interaction surface is depicted in Fig. 5-10. To avoid the numerous mathematical
complexities associated with the exact surface, several approximate techniques have been developed that relate
the response of a column in biaxial bending to its uniaxial resistance about each principal axis (Reference 5.5
summarizes a number of these approximate methods). A conservative estimate of the nominal axial load
strength can be obtained from the following (see ACI R10.3.6 and Fig. 5-11):
Mnx
Mb, Pb
Mx
My
Mny
Po
P
M
n
Figure 5-10 Biaxial Interaction Surface
ey
x
Pu
ex
y
y
x
Mux = Puey
Muy = Puex
Reinforcing bars not shown
Figure 5-11 Notation for Biaxial Loading

where Pni = nominal axial load strength for a column subjected to an axial load Pu at eccentricities ex and ey
Pnx = nominal axial load strength for a column subjected to an axial load Pu at eccentricity of ex only
(ey = 0)
Pny = nominal axial load strength for a column subjected to an axial load Pu at eccentricity ey only
(ex = 0)
Po = nominal axial load strength for a column subjected to an axial load Pu at eccentricity of zero (i.e.,
ex = ey = 0)
= 0.85 › (Ag
–Ast
) + fy
Ast
The above equation can be rearranged into the following form:
In design, Pu
φPni
where Pu
is the factored axial load acting at eccentricities ex
and ey
. This method is most
suitable when φPnx
and φPny
are greater than the corresponding balanced axial loads; this is usually the case for
typical building columns.
An iterative design process will be required when using this approximate equation for columns subjected to
biaxial loading. A trial section can be obtained from Figs. 5-18 through 5-25 with the factored axial load Pu
and
the total factored moment taken as Mu
= Mux
+ Muy
where Mux
= Pu
ex
and Muy
= Pu
ey
. The expression for φPni
can then be used to check if the section is adequate or not. Usually, only an adjustment in the amount of rein-
forcement will be required to obtain an adequate or more economical section.
5.5.3.1 Example: Simplified Design of a Column Subjected to Biaxial Loading
Determine the size and reinforcement for a corner column subjected to Pu
= 360 kips, Mux
= 50 ft-kips, and Muy
= 25 ft-kips.
(1) Trial section
From Fig. 5-20 with Pu
= 360 kips and Mu
= 50 + 25 = 75 ft-kips, select a 14 × 14 in. column with 4-
No.9 bars.
1
φPni
=
1
φPnx
+
1
φPny
−
1
φPo
φPni
=
1
1
φPnx
+
1
φPny
−
1
φPo
5-18

5-19
(2) Check the column using the approximate equation
For bending about the x-axis:
φPnx
= 455 kips for Muxux
= 50 ft-kips (see Fig. 5-20)
For bending about the y-axis:
φPny
= 464 kips for Muy
= 25 ft-kips (see Fig. 5-20)
φPo
= 0.65[0.85 × 4(142
– 4.0) + (60 × 4.0)] = 580 kips
φPni
= = 380 kips 360 kips OK
use a 14 × 14 in. column with 4-No.9 bars.
For comparison purposes, spColumn was used to check the adequacy of the 14 × 14 in. column with 4-No. 8
bars. Fig. 5-12 is the output from the program which is a plot of φMny
versus φMnx
for φPn
= Pu
= 360 kips (i.e.,
a horizontal slice through the interaction surface at φPn
= 360 kips). Point 1 represents the position of the
applied factored moments for this example. As can be seen from the figure, the section reinforced with 4-No.
8 bars is adequate to carry the applied load and moments. Fig. 5-13 is also output from the spColumn
program; this vertical slice through the interaction surface also reveals the adequacy of the section.
As expected, the approximate equation resulted in a more conservative amount of reinforcement (about 27%
greater than the amount from spColumn).
φPni
=
1
1
464
+
1
455
−
1
580
=

5-20
Figure 5-12 Moment Contour for 14 x 14 in. Column at φPn = 360 kips
My (k-ft)200
100
-100
-200
1
-200 -100 100 200
Mx (k-ft)
P = 360 kip

5-21
Figure 5-13 Interaction Diagram for 14 x 14 in. Column
P (kip)
(Pmax)
500
300
100
-100
-300
1
20 40 60 80 100
M (27°) (k-ft)
(Pmin)

5-22
* For a discussion of fixity of column bases, see PCI Design Handbook-Precast and Prestressed Concrete, 5th Ed.,
Precast/Prestressed Concrete Institute, Chicago, IL, 1999.
** The effective length factor k may be determined for a non-sway or sway frame using ACI R10.10 or using the simplified equations
which are also given in ACI R10.10.
5.6 COLUMN SLENDERNESS CONSIDERATIONS
5.6.1 Non-sway versus Sway Frames
The behavior of a column differs depending on whether it is a part of a sway or nonsway frame. Accordingly,
when designing columns, it is important to establish whether or not the building frame is nonsway. A column
may be assumed nonsway if located in a story in which the bracing elements (shear walls, shear trusses, or other
types of lateral bracing) have a such substantial lateral stiffness, to resist lateral movement of the story that the
resulting lateral deflection is not large enough to affect the column strength substantially. There is rarely a com-
pletely nonsway or a completely sway frame. Realistically, a column within a story can be considered nonsway
when horizontal displacements of the story do not significantly affect the moments in the column. ACI 10.10.5
gives criteria that can be used to determine if column located within a story is nonsway or sway. As a simpli-
fied approach, 10.10.1 permits the column to be considered braced against sidesway when the bracing elements
have a total stiffness, resisting the lateral movement of a story, of at least 12 times the gross stiffness of the
columns within the same story.
5.6.2 Minimum Sizing for Design Simplicity
Another important aspect to consider when designing columns is whether slenderness effects must be included
in the design (ACI 10.10). In general, design time can be greatly reduced if 1) the building frame is adequately
braced by shearwalls and 2) the columns are sized so that effects of slenderness may be neglected. The criteria
for the consideration of column slenderness, as prescribed in ACI 10.10, are summarized in Fig. 5-14. M2b
is
the larger factored end moment and M1b
is the smaller end moment; both moments, determined from an elastic
frame analysis, are due to loads that result in no appreciable side sway. The ratio M1b
/M2b
is positive if
the column is bent in single curvature, negative if it is bent in double curvature. For non-sway columns, the
effective length factor k = 1.0 (ACI 10.10.6.3).
In accordance with ACI 10.10.1, effects of slenderness may be neglected when non-sway columns are sized to
satisfy the following:
where ˜u
is the clear height between floor members and h is the column size. The above equation is valid for
columns that are bent in double curvature with approximately equal end moments. It can be used for the first
story columns provided the degree of fixity at the foundation is large enough.*
Table 5-4 gives the maximum
clear height ˜u
for a column size that would permit slenderness to be neglected.
For a sway column with a column-to-beam stiffness ratio ψ = 1 at both ends, the effects of slenderness may be
neglected with ˜u/h is less than 5, assuming k = 1.3 (see the alignment chart, in ACI R 10.10).**
u
h
≤ 12

If the beam stiffness is reduced to one-fifth of the column stiffness at each end, then k = 2.2; consequently,
slenderness effects need not be considered as long as ˜u/h is less than 3. As can be seen from these two
examples, beam stiffnesses at the top and the bottom of a column in a structure where sidesway is not prevented
will have a significant influence on the degree of slenderness of the column.
Due to the complexities involved, the design of slender columns is not considered in this book. For a
comprehensive discussion of this topic, the reader is referred to Chapter 11 of Reference 5.5.
5-23
Zone 1
Zone 2
Zone 3
Single
curvature
Double
curvature
M1/M2
k˜u
r
1.0
-1.0
Zone 1: Neglect slenderness, sway and non-sway frames
Zone 2: Neglect slenderness, non-sway frames
Zone 3: Consider slenderness, moment magnification method (ACI 10.10.6 and ACI 10.10 .7)
22 34
40
-0.5
Table 5-4 Maximum Story Heights to Neglect Slenderness—Non-sway Columns
Column size
h (in.)
Maximum clear height
u (ft)
10 10
12 12
14 14
16 16
18 18
20 20
22 22
24 24
Figure 5-14 Consideration of Column Slenderness

5.7 PROCEDURE FOR SIMPLIFIED COLUMN DESIGN
The following procedure is suggested for design of a multistory column stack using the simplifications and
column design charts presented in this chapter. For sway frames with non-slender columns, both gravity and
wind loads must be considered in the design. Figs. 5-18 through 5-25 can be used to determine the required
reinforcement. For non-sway frames with shearwalls resisting the lateral loads and the columns sized so that
slenderness may be neglected, only gravity loads need to be considered; the reinforcement can be selected for
Figs. 5-18 through 5-25 as well.
STEP (1) LOAD DATA
(a) Gravity Loads:
Determine factored loads Pu
for each floor of the column stack being considered. Include a service dead
load of 4 kips per floor for column weight. Determine column moments due to gravity loads. For inte-
rior columns supporting a two-way floor system, maximum column moments may be computed by
ACI Eq. (13-7) (see Chapter 4, Section 4.5). Otherwise, a general analysis is required.
(b) Lateral Loads:
Determine axial loads and moments from the lateral loads for the column stack being considered.
STEP (2) LOAD COMBINATIONS
For gravity (dead + live) plus lateral loading, ACI 9.2 specifies five load combinations that need to be
considered (Table 2-6).
STEP (3) COLUMN SIZE AND REINFORCEMENT
Determine an initial column size based on the factored axial load Pu in the first story using Fig. 5-2,
and use this size for the full height of building. Note that the dimensions of the column may be preset
by architectural (or other) requirements. Once a column size has been established, it should be deter-
mined if slenderness effects need to be considered (see Section 5.6). For columns with slenderness
ratios larger than the limits given in ACI 10.10, it may be advantageous to increase the column size (if
possible) so that slenderness effects may be neglected.
As noted earlier, for nonslender columns, Figs. 5-18 through 5-25 may be used to select the required
amount of reinforcement for a given Pu and Mu. Ideally, a column with a reinforcement ratio in the
range of 1% to 2% will result in maximum economy. Depending on the total number of stories,
differences in story heights, and magnitudes of lateral loads, 4% to 6% reinforcement may be required
in the first story columns. If the column bars are to be lap spliced, the percentage of reinforcement
should usually not exceed 4% (ACI R10.9.1). For overall economy, the amount of reinforcement can
be decreased at the upper levels of the building. In taller buildings, the concrete strength is usually
varied along the building height as well, with the largest › used in the lower level(s).
5-24

5.8 EXAMPLES: SIMPLIFIED DESIGN FOR COLUMNS
The following examples illustrate the simplified methods presented in this chapter.
5.8.1 Example: Design of an Interior Column Stack for Building #2 Alternate (1)—
Slab and Column Framing Without Structural Walls (Sway Frame)
› = 4000 psi (carbonate aggregate)
fy
= 60,000 psi
(1) LOAD DATA
Roof: LL = 20 psf Floors: LL = 50 psf
DL = 122 psf DL = 136 psf (8.5 in. slab)
Calculations for the first story interior column are as follows:
(a) Total factored load:*
Factored axial loads due to gravity are summarized in Table 5-5.
(b) Factored moments:
gravity loads:
The moment due to dead load is small. Only moments due to live loads and wind loads will
be considered.
5-25
Table 5-5 Interior Column Gravity Load Summary for Building #2—Alternate (1)
Floor Dead
Load
(psf)
Live
Load
(psf)
Tributary
Area
(sq ft)
Influence
Area
RM Reduced
Live Load
(psf)
Cumulative
Dead Load
(kips)
Cumulative
Live Load
(kips)
Cumulative
Factored
Load
ACI-Eq. (9-2)
(kips)
5th (roof) 122 20 480 — 1 20.0 63 9.6 80
4th 136 50 480 1920 0.59 29.5 132 23.8 186
3rd 136 50 480 3840 0.49 24.5 201 35.5 288
2nd 136 50 480 5760 0.45 22.5 270 46.3 388
1st 136 50 480 7680 0.42 21.0 340 56.4 487
* Axial load from wind loads is zero (see Fig. 2-15).

Live load moment = 0.035 qLu
˜2
˜n
2
= 0.035(0.05)(24)(18.832) = 14.9 ft-kips ACI Eq. (13.4)
Portion of live load moment to first story column = 14.9 = 6.6 ft-kip
(2) LOAD COMBINATIONS
For the 1st story column:
gravity loads: Pu
= 487 kips ACI Eq. (9-2)
Mu
= 1.6(6.6) = 10.6 ft-kips
gravity loads + wind loads: Pu
= 1.2(340) + 1.6 (9.6) = 423 kips ACI Eq. (9-3)
Mu
= 1.6(6.6) + 0.8(68.75) = 65.6 ft-kips
or
Pu
= 1.2(340) + 0.5(56.4) = 436 kips ACI Eq. (9-4)
Mu
= 0.50(6.6) + 1.6(68.75) = 113 ft-kips
or
Pu
= 0.9(340) = 306 kips ACI Eq. (9-6)
Mu
= 1.6(68.75) = 110 ft-kips
Factored loads and moments, and load combinations, for the 2nd through 5th story columns are calculated in
a similar manner, and are summarized in Table 5-6.
12
12 +15
⎡
⎣⎢
⎤
⎦⎥
5-26
Table 5-6 Interior Column Load Summary for Building #2, Alternate (1)
Floor ACI Eq. (9-2) ACI Eq. (9-3) ACI Eq. (9-4) ACI Eq. (9-6)
Pu Mu Pu Mu Pu Mu Pu Mu
5th (roof) 80 10 90 15 80 13 56 10
4th 186 12 174 25 170 34 119 30
3rd 288 12 257 34 259 54 181 50
2nd 388 13 340 44 348 73 243 68
1st 487 11 423 65 436 113 306 110

(3) COLUMN SIZE AND REINFORCEMENT
With Pu
= 487 kips, try a 16 × 16 in. column with 1% reinforcement (see Fig. 5-2). Check for fire resistance:
From Table 10-2, for a fire resistance rating of 2 hours, minimum column dimension = 10 in. 16 in. O.K.
Determine if the columns are slender.
As noted above, a column in a sway frame is slender if k˜u
/r ≥ 22. In lieu of determining an “exact” value,
estimate k to be 1.2 (a value of k less than 1.2 is usually not realistic for columns in a sway frame.
For a 1st story column:
For the 2nd through 5th story columns:
Therefore, slenderness must be considered for the entire column stack. To neglect slenderness effects, the size
of the column h would have to be:
Obviously, this column would not be practical for a building of the size considered. Reference 5.4 or 5.5 can
be used to determine the required reinforcement for the 16 × 16 in. column, including slenderness effects.
Figure 5.15 shows the results from spColumn for an interior 1st story column, including slenderness effects.
Thirty five percent of the gross moment of inertia of the slab column strip and seventy percent of the gross
moment of inertia of the column section were used to account for the cracked cross section.* It was assumed
that the column was fixed at the foundation; appropriate modifications can be made if this assumption is not
true, based on the actual footing size and soil conditions. Points 2, 4, 6, 8 and 10 correspond to the
load combinations given in ACI Eq. (9-1), (9-2), (9-3), (9-4) and (9-6) respectively. As can be seen from the
figure, 8-No.9 bars are required at the 1st floor. The amount of reinforcement can decrease at higher
elevations in the column stack.
Check for fire resistance: From Table 10-6, for a fire resistance rating of 4 hours or less, the required cover to
the main longitudinal reinforcement = 1.5 in. provided cover = 1.875 in. O.K.
1.2 15 ×12( )− 8.5⎡⎣ ⎤⎦
0.3h
22 → h 31.2in.
k u
r
=
1.2 12 ×12( )− 8.5⎡⎣ ⎤⎦
0.3 16( )
= 34 22
k u
r
=
1.2 15 ×12( )− 8.5⎡⎣ ⎤⎦
0.3 16( )
= 43 22
5-27
* The moments of inertia of the flexural and compression members are required in order to compute the effective length factor k of
the column. ACI R10.10 recommends using a value of 0.35 Ig for flexural members (to account for the effect of cracking and rein-
forcement on relative stiffness) and 0.70Ig for compression members when computing the relative stiffness at each end of the com-
pression member, where Ig is the gross moment of inertia of the section.

5-28
Figure 5-15 Interaction Diagram for First Story Interior Column, Building #2,
Alternate (1), Including Slenderness
P (kip)
(Pmax)
900
700
500
300
100
-100
-300
-500
3
100 200 300
Mx (k-ft)
(Pmin)
2 23
14
16 9
1826

5-29
5.8.2 Example: Design of an Interior Column Stack for Building #2 Alternate (2) –
Slab and Column Framing with Structural Walls (Non-sway Frame)
= 4000 psi (carbonate aggregate)
fy
= 60,000 psi
For the Alternate (2) framing, columns are designed for gravity loading only; the structural walls are designed
to resist total wind loading.
(1) LOAD DATA
DL = 122 psf DL = 142 psf (9 in. slab)
Calculations for the first story interior column are as follows:
(a) Total factored load (see Table 5-7):
(b) Factored gravity load moment:
Mu
= 0.035 qLu
˜2
˜n
2 = 0.035(1.6 x 0.05)(24)(18.832
) = 23.8 ft-kips
portion of Mu to 1st story column = 24 (12/27) = 10.6 ft-kips
Similar calculations can be performed for the other floors.
The applicable load combination for each floor is summarized in Table 5-8. Note that only ACI Eq. (9-2) needs
to be considered for columns in a non-sway frame.
ʹfc
Table 5-7 Interior Column Gravity Load Summary for Building #2, Alternate (2)
Floor Dead
Load
(psf)
Live
Load
(psf)
Tributary
Area
(sq ft)
Influence
Area
(sq ft)
RM Reduced
Live Load
(psf)
Cumulative
Dead Load
(kips)
Cumulative
Live Load
(kips)
Cumulative
Factored
Load ACI-
Eq. (9-2)
(kips)
5th
(roof) 122 20 480 -- 1 20 63 9.6 80
4th
142 50 480 1920 0.59 29.5 135 23.76 189
3rd
142 50 480 3840 0.49 24.5 207 35.52 295
2nd
142 50 480 5760 0.45 22.5 279 46.32 398
1st
142 50 480 7680 0.42 21 351 56.4 501

5-30
With Pu
= 501 kips, try a 16 × 16 in. column with 1% reinforcement (see Fig. 5-2).
Check for fire resistance: From Table 10-2, for a fire resistance rating of 2 hours, minimum column dimension
= 10 in. 16 in. O.K.
Using Table 5-4, for a 16 in. column, the maximum clear story height to neglect slenderness is 18.67 ft. Since
the actual clear story heights are less than this value, slenderness need not be considered for the entire column
stack.
• 1st story columns:
Pu
= 501 kips, Mu
= 10.6 ft-kips
From Fig. 5-21, use 4-No. 8 bars (ρg
= 1.23%)
• 2nd through 5th story columns:
Using 4-No. 8 bars for the entire column stack would not be economical. ACI 10.8.4 may be used so that the
amount of reinforcement at the upper levels may be decreased. The required area of steel at each floor can be
obtained from the following:
Required Ast
= (area of 4-No. 8 bars)
where 546 kips is φPn for the 16 × 16 in. column reinforced with 4-No. 8 bars. It is important to note that ρg
should never be taken less than 0.5% (ACI 10.8.4). The required reinforcement for the column stack is
summarized in Table 5-9.
Pu at floor level
546 kips
Table 5-8 Interior Column Load Summary for Building #2, Alternate (2)
Floor ACI Eq. (9-1)
Pu
(kips)
Mu
(ft-kips)
5th
80 9.5
4th
189 11.9
3rd
295 11.9
2nd
398 13.2
1st
501 10.6

Check for fire resistance: From Table 10-6, for a fire resistance rating of 4 hours or less, the required cover to
the main longitudinal reinforcement = 1.5 in. provided cover. O.K.
Column ties and spacing can be selected from Table 5-3.
5.8.3 Example: Design of an Edge Column Stack (E-W Column Line) for Building
#1—3-story Pan Joist Construction (Sway Frame)
= 4000 psi (carbonate aggregate)
fy
= 60,000 psi
Required fire resistance rating = 1 hour (2 hours for columns supporting Alternate (2) floors).
(1) LOAD DATA
DL = 105 psf DL = 130 psf
Calculations for the first story column are as follows:
(a) Total factored load (see Table 5-10):
ʹfc
5-31
Floor Required Ast
(in2
)
Required g
(%)
Reinforcement
( g%)
5th (roof) 0.52 0.5 4- #6 (0.69)
4th 1.16 0.5 4- #6 (0.69)
3rd 1.76 0.69 4- #6 (0.69)
2nd 2 0.78 4- #7 (0.94)
1st 3.16 1.23 4- #8 (1.23)
Table 5-9 Reinforcement for Interior
Column of Building #2, Alternat (2)
Table 5-10 Edge Column Gravity Load Summary for Building #1
Floor Dead
Load
(psf)
Live
Load
(psf)
Tributary
Area
(sq ft)
Influence
Area
RM Reduced
Live Load
(psf)
Cumulative
Dead Load
(kips)
Cumulative
Live Load
(kips)
Cumulative
Factored
Load ACI
Eq. (9-2)
(kips)
3rd
(roof) 105 12 450 -- 1 12 51 5.4 64
2nd
130 60 450 1800 0.6 36 114 21.6 165
1st
130 60 450 3600 0.5 30 176 35.1 262

(b) Factored moments in 1st story edge columns:
gravity loads: Mu
= 327.6 ft-kips (see Section 3.8.3 – Step (2), Mu
@ exterior columns) portion of Mu
to 1st
story column = 327.6/2 = 164 ft-kips
wind loads (see Fig. 2-13):
P = 10.91 kips
M = 61.53 ft-kips
For the 1st story column:
gravity loads:
Pu
= 268 kips ACI Eq. (9-2)
Mu
= 164 ft-kips
Pu
= 1.2(176) + 0.8(10.91) = 220 kips ACI Eq. (9-3)
Mu
= 1.2(99.5) + 0.8(61.53) = 169 ft-kips
or
Pu
= 1.2(176) + 0.50(35.1) + 1.6(10.91) = 247 kips ACI Eq. (9-4)
Mu
= 1.2(99.5) + 0.50(27.7) + 1.6(61.53) = 232 ft-kips
or
Pu = 0.9(176) + 1.6(10.91) = 176 kips ACI Eq. (9-6)
Mu = 0.9(99.5) + 1.6(61.53) = 188 ft-kips
Factored loads and moments, and load combinations, for the 2nd and 3rd story columns are calculated in a sim-
ilar manner, and are summarized in Table 5-11.
5-32

For edge columns, initial selection of column size can be determined by referring directly to the column design
charts and selecting an initial size based on required moment strength. For largest Mu
= 233 kips, try a 16 × 16
in. column (see Fig. 5-21).
Check for fire resistance: From Table 10-2, for fire resistance ratings of 1 hour and 2 hours, minimum column
dimensions of 8 in. and 10 in., respectively, are both less than 16 in. O.K.
Using k = 1.2, slenderness ratios for all columns:
Thus, all of the columns are slender. To neglect slenderness effects, the size of the column would have to be:
This column would probably not be practical for a building of the size considered.
Fig. 5-16 shows the results from spColumn for a first story edge column, including slenderness effects. Thirty
five percent of the gross moment of inertia was used for the 36 × 19.5 in. column-line beam and seventy
percent of the gross moment of inertia of the column cross section to account for cracking. The column was
assumed fixed at the foundation. As can be seen from the figure, 8-No. 11 bars are required in this case.
Check for fire resistance: From Table 10-6, for fire resistance ratings of 4 hours or less, required cover to main
longitudinal reinforcement is 1.5 in. provided cover = 1.875 in. O.K.
1.2 13×12( )−19.5⎡⎣ ⎤⎦
0.3h
22 → h 24.8in.
k u
r
=
1.2 13×12( )−19.5⎡⎣ ⎤⎦
0.3 16( )
= 34 22
5-33
Floor ACI Eq. (9-2) ACI Eq. (9-3) ACI Eq. (9-4) ACI Eq. (9-6)
P M Pu Mu Pu Mu Pu Mu
3rd 64 202 71 233 66 223 48 166
2nd 165 164 140 150 154 194 109 150
1st 262 164 220 169 247 232 176 188
Table 5-11 Edge Column Load Summary for Building #1

5-34
Figure 5-16 Interaction Diagram for First Story Edge Column, Building #1, Including Slenderness
P (kip)
(Pmax)
1200
800
400
-400
-800
11
100 200 300
Mx (k-ft)
(Pmin)
424
14
16
9
1826
7

5-35
5.9 COLUMN SHEAR STRENGTH
Columns in sway frames are required to resist the shear forces from lateral loads. For members subjected to
axial compression, the concrete shear strength φVc is given in ACI Eq. (11-4). Fig. 5-17 can be used to obtain
this quantity for the square column sizes shown. The largest bar size from the corresponding column design
charts of Figs. 5-18 through 5-25 were used to compute φVc (for example, for a 16 × 16 in. column, the largest
bar size in Fig. 5-21 is No. 11).
ACI Eq. (9-6) should be used to check column shear strength:
U = 0.9D + 1.6W
Nu
= Pu
= 0.9D
Vu
= 1.6W
If Vu
is greater than φVc, spacing of column ties can be reduced to provide additional shear strength φVs. Using
the three standard spacings given in Chapter 3, Section 3.6, the values of φVs given in Table 5-12 may be used
to increase column shear strength.
For low-rise buildings, column shear strength φVc will usually be more than adequate to resist the shear forces
from wind loads.
5.9.1 Example: Design for Column Shear Strength
Check shear strength for the 1st floor interior columns of Building No. 2, Alternate (1) – slab and column
framing without structural walls. For wind in the N-S direction, V = 9.17 kips (see Fig. 2-17).
Nu
= Pu
= 306 kips (see Example 5.7.1)
Vu
= 1.6(9.17) = 14.67 kips
From Fig. 5-17, for a 16 ×16 in. column with Nu
= 306 kips:
φVc Х 32 kips 16.42 kips O.K.
Column shear strength is adequate. With No. 10 column bars, use No. 3 column ties at 16 in. on center (least
column dimension governs; see Table 5-3).
Table 5-12 Shear Strength Provided by Column Ties
Tie
Spacing
φVs - #3 ties* φVs - #4 ties*
d/2 19 kips 35 kips
d/3 29 kips 54 kips
d/4 40 kips 71 kips
*2 legs, Grade 60 bars

5-36
0
10
20
30
40
50
60
70
80
0 100 200 300 400 500 600
Nu = Pu, kips
φVc,kips
24 x 24
22 x 22
20 x 20
18 x 18
16 x 16
14 x 14
12 x 12
10 x 10
fc' = 4000 psi
φ = 0.75
φVc = 2φ 1+
Nu
2000Ag
fc' bwd
Figure 5-17 Column Shear Strength, φVc

5-37
0.00
50.00
100.00
150.00
200.00
250.00
300.00
0 10 20 30 40 50 60
Mu = φMn, ft-kips
Pu=φPn,kips
4-#8 (3.16)
4-#6 (1.76)
4-#5 (1.24)
10 10 in. column
fc' = 4000 psi
fy = 60 ksi
4-#9 (4.00)
4-#7 (2.40)
x
Figure 5-18 – 10 x 10 in. Column Design Chart

5-38
0.00
100.00
200.00
300.00
400.00
500.00
0 20 40 60 80 100 120
Mu = φMn, ft-kips
Pu=φPn,kips
12 12 in. column
fc' = 4000 psi
fy = 60 ksi
4-#9 (2.78)
4-#8 (2.19)
4-#7 (1.67)
4-#6 (1.22)
4-#11 (4.33)
4-#10 (3.53)
8-#9 (5.56)
x

5-39
0.00
100.00
200.00
300.00
400.00
500.00
600.00
700.00
800.00
0 50 100 150 200 250
Mu = φMn, ft-kips
14 14 in. column
fc' = 4000 psi
fy = 60 ksi
4-#9 (2.04)
4-#8 (1.61)
4-#7 (1.22)
8-#9 (4.08)
4-#11 (3.18)
4-#10 (2.59)
8-#11 (6.37)
8-#10 (5.18)
Pu=φPn,kips
x

5-40
0.00
200.00
400.00
600.00
800.00
1000.00
0 50 100 150 200 250 300 350 400 450 500
Mu = φMn, ft-kips
16 16 in. column
fc
' = 4000 psi
fy = 60 ksi
4-#9 (1.56)
4-#8 (1.23)
8-#9 (3.13)
4-#11 (2.44)
4-#10 (1.98)
8-#11 (4.88)
8-#10 (3.97)
Pu=φPn,kips
12-#11 (7.31)
12-#10 (5.95)
x

5-41
0.00
200.00
400.00
600.00
800.00
1000.00
1200.00
0 50 100 150 200 250 300 350 400 450 500
Mu = φMn, ft-kips
18 18 in. column
fc' = 4000 psi
fy = 60 ksi
4-#9 (1.23)
8-#9 (2.47)
4-#11 (1.93)
4-#10 (1.57)
8-#11 (3.85)
8-#10 (3.14)
Pu=φPn,kips
12-#11 (5.78)
12-#10 (4.70)
16-#10 (6.27)
x

5-42
0.00
200.00
400.00
600.00
800.00
1000.00
1200.00
1400.00
1600.00
0 100 200 300 400 500 600 700
Mu = φMn, ft-kips
20 20 in. column
fc' = 4000 psi
fy = 60 ksi
4-#9 (1.00)
8-#9 (2.00)
4-#11 (1.56)
4-#10 (1.27)
8-#11 (3.12)
8-#10 (2.54)
Pu=φPn,kips
12-#11 (4.68)
12-#10 (3.81)
16-#10 (5.08)
16-#11 (6.24)
x

5-43
0.00
200.00
400.00
600.00
800.00
1000.00
1200.00
1400.00
1600.00
0 100 200 300 400 500 600 700 800
Mu = φMn, ft-kips
22 22 in. column
fc' = 4000 psi
fy = 60 ksi
8-#9 (1.65)
4-#11 (1.29)
4-#10 (1.05)
8-#11 (2.58)
8-#10 (2.10)
Pu=φPn,kips
12-#11 (3.87)
12-#10 (3.15)
16-#10 (4.20)
16-#11 (5.16) x

5-44
0.00
200.00
400.00
600.00
800.00
1000.00
1200.00
1400.00
1600.00
1800.00
0 100 200 300 400 500 600 700 800 900
Mu = φMn, ft-kips
P
u
φ
P
n
24 24 in. column
fc' = 4000 psi
fy = 60 ksi
8-#9 (1.39)
4-#11 (1.08)
8-#11 (2.17)
8-#10 (1.76)
Pu=φPn,kips
12-#11 (3.25)
12-#10 (2.65)
16-#10 (3.53)
16-#11 (4.33)
x

5-45
References
5.1 Strength Design of Reinforced Concrete Columns, Portland Cement Association, Skokie, Illinois,
EB009.02D, 1977, 49 pp.
5.2 Design Handbook in Accordance with the Strength Design Method of ACI 318-95: Vol. 2–Columns,
SP17A(90).CT93, American Concrete Institute, Detroit, Michigan, 1997, 222 pp.
5.3 CRSI Handbook, 9th Edition, Concrete Reinforcing Steel Institute, Schaumburg, Illinois, 2002.
5.4 spColumn, Design and investigation of reinforced concrete column sections, Structure Point, Skokie,
Illinois, 2010.
5.5 Notes on ACI 318-08, 8th Edition, EB708, Portland Cement Association, Skokie, Illinois 2008.

6-1
Chapter 6
Simplified Design for Structural Walls
6.1 INTRODUCTION
For buildings in the low to moderate height range, frame action alone is usually sufficient to provide adequate
resistance to lateral loads. Whether directly considered or not, nonstructural walls and partitions can also add
to the total rigidity of a building and provide reserve capacity against lateral loads.
Structural walls or shearwalls are extremely important members in high-rise buildings. If unaided by walls,
high-rise frames often could not be efficiently designed to satisfy strength requirements or to be within
acceptable lateral drift limits. Since frame buildings depend primarily on the rigidity of member connections
(slab-column or beam-column) for their resistance to lateral loads, they tend to be uneconomical beyond
a certain height range (11-14 stories in regions of high to moderate seismicity, 15-20 stories elsewhere).
To improve overall economy, structural walls are usually required in taller buildings.
If structural walls are to be incorporated into the framing system, a tentative decision needs to be made at the
conceptual design stage concerning their location in plan. Most multi-story buildings are constructed with a
central core area. The core usually contains, among other things, elevator hoistways, plumbing and HVAC
shafts, and possibly exit stairs. In addition, there may be other exit stairs at one or more locations remote from
the core area. All of these involve openings in floors, which are generally required by building codes to be
enclosed with walls having a fire resistance rating of one hour or two hours, depending on the number of
stories connected. In general, it is possible to use such walls for structural purposes.
If at all possible, the structural walls should be located within the plan of the building so that the center of
rigidity of the walls coincides with the line of action of the resultant wind loads or center of mass for seismic
design (see Chapter 11). This will prevent torsional effects on the structure. Since concrete floor systems act
as rigid horizontal diaphragms, they distribute the lateral loads to the vertical framing elements in proportion
to their rigidities. The structural walls significantly stiffen the structure and reduce the amount of lateral drift.
This is especially true when shearwalls are used with a flat plate floor system.
6.2 FRAME-WALL INTERACTION
The analysis and design of the structural system for a building frame of moderate height can be simplified if
the structural walls are sized to carry the entire lateral load. Members of the frame (columns and beams or
slabs) can then be proportioned to resist the gravity loads only. Neglecting frame-wall interaction for buildings
of moderate size and height will result in reasonable member sizes and overall costs. When the walls stiffness
is much higher than the stiffness of the columns in a given direction within a story, the frame takes only a small
portion of the lateral loads. Thus, for low-rise buildings, neglecting the contribution of frame action in resisting
lateral loads and assigning the total lateral load resistance to walls is an entirely reasonable assumption. In

contrast, frame-wall interaction must be considered for high-rise structures where the walls have a significant
effect on the frame: in the upper stories, the frame must resist more than 100% of the story shears caused by
the wind loads. Thus, neglecting frame-wall interaction would not be conservative at these levels. Clearly, a
more economical high-rise structure will be obtained when frame-wall interaction is considered.
With adequate wall bracing, the frame can be considered non-sway for column design. Slenderness effects
can usually be neglected, except for very slender columns. Consideration of slenderness effects for sway and
non-sway columns is discussed in Chapter 5, Section 5.6.
6.3 WALL SIZING FOR LATERAL BRACING
The size of openings required for stairwells and elevators will usually dictate minimum wall plan layouts.
From a practical standpoint, a minimum thickness of 6 in. will be required for a wall with a single layer of
reinforcement, and 10 in. for a wall with a double layer (ACI 14.3.4). While fire resistance requirements will
seldom govern wall thickness, the building code requirements should not be overlooked. See Chapter 10 for
design considerations for fire resistance. The above requirements will, in most cases, provide stiff enough walls
so that the frame can be considered non-sway.
The designer has to distinguish between sway and non-sway frames. This can be done by comparing the total
lateral stiffness of the columns in a story to that of the bracing elements. A compression member may be
considered non-sway if it is located in a story in which the bracing elements (shearwalls) have such substantial
lateral stiffness to resist the lateral deflection of the story that any resulting deflection is not large enough to affect
the column strength substantially. A simple criterion is given in ACI 10.10.1 to establish whether structural walls
provide sufficient lateral bracing to qualify the frame as braced: The structural walls must have a total stiffness at
least equal to twelve times the sum of the stiffnesses of all the columns in a given direction within a story:
⌺EI(walls)
≥ ⌺12EI(columns)
The above criterion can be used to size the structural walls within the range of structures covered in this
publication so that the frame can be considered non-sway.
6.3.1 Example: Wall Sizing for Non-Sway Condition
Using the approximate criteria given in ACI 10.10.1, size the structural walls for Alternate (2) of Building #2
(5 story flat plate)*. In general, both the N-S and E-W directions must be considered. The E-W direction will
be considered in this example since the moment of inertia of the walls will be less in this direction. The plan
of Building #2 is shown in Fig. 6-1.
Required fire resistance rating of exit stair enclosure walls = 2 hours
For interior columns: I = (1
/12
)(164
) = 5461 in.4
For edge columns: I = (1
/12)(124
) 1728 in.4
I(columns) = 8(5461) + 12(1728) = 64,424 in.4
12I(columns) = 773,088 in.4
* The 5-story flat plate frame of Building #2 is certainly within the lower height range for structural wall consideration.
Both architectural and economic considerations need to be evaluated to effectively conclude if structural walls need to be
included in low-to-moderate height buildings.
6-2

Try an 8 in. wall thickness. To accommodate openings required for stairwells, provide 8 ft flanges as shown in
Fig. 6-2.
From Table 10-1, for a fire resistance rating of 2 hours, required wall thickness = 4.6 in. ≤ 8 in. O.K.
E-W direction
Ag
= (248 ϫ 8) + (88 ϫ 8 ϫ 2) = 1984 + 1408 = 3392 in.2
x = [(1984 ϫ 4) + (1408 ϫ 52)]/3392 = 23.9 in.
Iy
= [(248 ϫ 83/12) + (1984 ϫ 19.92)] + [2(8 ϫ 883/12) + (1408 ϫ 28.12)] = 2,816,665 in.4
For two walls: I(walls)
= 2(2,816,665) = 5,663,330 in.4 773,088 in.4
6-3
Chapter 6 • Simplified Design for Structural Walls
N
16 x 16 typ.
12 x 12 typ.
Exit stair
Enclosure
walls
Exit stair
Enclosure
walls
Figure 6-1 Plan of Building #2, Alternate (2)
8 typ.
y
c.g.
x
x=23.9
96
(8-0)
248
(20-8)
Figure 6-2 Plan View of Shearwall

Therefore, the frame can be considered non-sway for column design. Since the wall segments in the E-W
direction provide most of the stiffness in this direction, the 8 ft length provided for the stairwell enclosure is
more than adequate.
6.4 DESIGN FOR SHEAR
Design for horizontal shear forces (in the plane of the wall) can be critical for structural walls with small
height-to-length ratios (i.e., walls in low-rise buildings). Special provisions for walls are given in ACI 11.9. In
addition to shear, the flexural strength of the wall must also be considered (see Section 6.5).
Walls with minimum amounts of vertical and horizontal reinforcement are usually the most economical. If
much more than the minimum amount of reinforcement is required to resist the factored shear forces, a change
in wall size (length or thickness) should be considered. The amounts of vertical and horizontal reinforcement
required for shear depend on the magnitude of the factored shear force, Vu
. Table 6-1 summarizes the amounts
of vertical and horizontal reinforcement required for shear for structural walls.
Table 6-1 Shear Reinforcement for Structural Walls
(1) When the factored shear force is less than or equal to one-half the shear strength provided by concrete
(Vu ≤ φVc/2), minimum wall reinforcement according to ACI 14.3 must be provided. For walls subjected
to axial compressive forces, φVc may be taken as hd for normal weight concrete where h is the
thickness of the wall, d = 0.8˜w (ACI 11.9.4), and ˜w is the length of the wall (ACI 11.9.5). Suggested
vertical and horizontal reinforcement for this situation is given in Table 6-2.
(2) When the design shear force is more than one-half the shear strength provided by concrete (Vu φVc/2),
minimum shear reinforcement according to ACI 11.9.9 must be provided. Suggested reinforcement (both
vertical and horizontal) for this situation is given in Table 6-3.
(3) When the design shear force exceeds the concrete shear strength (Vu φVc), horizontal shear
reinforcement must be provided according to ACI Eq. (11-29). Note that the vertical and horizontal
reinforcement must not be less than that given in Table 6-3.
φ2λ ʹfc
6-4
Vu Horizontal Shear Reinforcement Vertical Shear Reinforcement
Vu ␾Vc/2 ρt = 0.0020 for #5 and smaller ρ = 0.0012 for #5 and smaller
ρt = 0.002 5 for other bars ρ = 0.0015 for other bars
␾Vc/2 V u ␾Vc ρt = 0.0025 ρ = 0.0025
Vu ␾Vc ␾Vs = ␾Avfyd/S2 ρ = 0.0025 + 0.5[2.5- ]( ρ)/(h ww t - 0.0025)
␾Vs + ␾Vc = Vu
␾Vs + ␾Vc 10␾ )8(0.hf w
'
c

Table 6-2 Minimum Wall Reinforcement (Vu
≤ φVc
/2)
Table 6-3 Minimum Wall Reinforcement (φVc
/2 ≤ Vu
≤ φVc
)
Using the same approach as in Section 3.6 for beams, design for required horizontal shear reinforcement in
walls when Vu
φVc
can be simplified by obtaining specific values for the design shear strength φVs
provided
by the horizontal reinforcement. As noted above, ACI Eq. (11-29) must be used to obtain φVs
:
where Av
is the total area of the horizontal shear reinforcement within a distance s2
, φ= 0.75. fy
= 60,000 psi,
and d = 0.8˜w (ACI 11.10.4). For a wall reinforced with No. 4 bars at 12 in. in a single layer, φVs becomes:
φVs
= 0.75 ϫ 0.20 ϫ 60 ϫ (0.8 ϫ 12˜w
)/12 = 7.2˜w
kips
where ˜w is the horizontal length of wall in feet.
Table 6-4 gives values of φVs
per foot length of wall based on various horizontal bar sizes and spacings.
6-5
Vertical Horizontal
Wall Thickness
h (in.)
Minimum As
a
(in.2
/ft)
Suggested
Reinforcement
Minimum As
b
(in.2
/ft)
Suggested
Reinforcement
6 0.09 #3 @ 15 0.14 #4 @ 16
8 0.12 #3 @ 11 0.19 #4 @ 12
10 0.14 #4 @ 16 0.24 #5 @ 15
12 0.17 #3 @ 15c
0.29 #4 @ 16c
a
Minimum As/ft of wall = 0.0012(12)h = 0.0144h for #5 bars and less (ACI 14.3.2)
b
Minimum As/ft of wall = 0.0020(12)h = 0.0240h for #5 bars and less (ACI 14.3.3)
c
Two layers of reinforcement are required (ACI 14.3.4)
Vertical and Horizontal
Wall Thickness
h (in.)
Minimum As
a
(in.2
/ft)
Suggested
Reinforcement
6 0.18 #4 @ 13
8 0.24 #4 @ 10
10 0.30 #5 @ 12
12 0.36 #4 @ 13b
a
Minimum As/ft of wall = 0.0025(12)h = 0.03h (ACI 11.10.9)
b
Two layers of reinforcement are required (ACI 14.3.4)
φVs
= φ
Avfy d
s2

Table 6-4 Shear Strength φVs
Provided by Horizontal Shear Reinforcement*
6-6
Vs (kips/ft length of wall)
Bars Spacing
S2 (in.) #3 #4 #5 #6
6 7.9 14.4 22.3 31.7
7 6.8 12.3 19.1 27.2
8 5.9 10.8 16.7 23.8
9 5.3 9.6 14.9 21.1
10 4.8 8.6 13.4 19.0
11 4.3 7.9 12.2 17.3
12 4.0 7.2 11.2 15.8
13 3.7 6.6 10.3 14.6
14 3.4 6.2 9.6 13.6
15 3.2 5.8 8.9 12.7
16 3.0 5.4 8.4 11.9
17 2.8 5.1 7.9 11.2
18 2.6 4.8 7.4 10.6
Values of Vs are for walls with a single layer of reinforcement.
Tabulated values can be doubled for walls with two layers
hw
s1
s2
˜w
ρ
ρt
*
Tabulated values should be doubled for walls with two layers

Table 6-5 gives values of and limiting values of ,
both expressed in kips per foot length of wall.
Table 6-5 Design Values of φVc
and Maximum Allowable φVn
The required amount of vertical shear reinforcement is given by ACI Eq. (11-30):
ρ˜
= 0.0025 + 0.5(2.5 – hw
/˜w
)(ρt
– 0.0025)
where hw
= total height of wall
ρ˜
= Avn
/s1
h
ρt
= Avh
/s2
h
When the wall height-to-length ratio hw
/˜w
is less than 0.5, the amount of vertical reinforcement is equal to the
amount of horizontal reinforcement (ACI 11.9.9.4).
6.4.1 Example 1: Design for Shear
To illustrate the simplified methods described above, determine the required shear reinforcement for the wall
shown in Fig. 6-3. The service shear force from the wind loading is 160 kips. Assume total height of wall from
base to top is 20 ft.
Figure 6-3 Plan View of Shearwall
φVn
= φVc
+ φVs
= φ10 ʹfc
h 0.8 w( )φVc
= φ2 ʹfc
h 0.8 w( )
6-7
Wall
Thickness
h (in.)
Vc
(kips/ft length of wall)
Max. Vn
(kips/ft length of wall)
6 5.5 27.3
8 7.3 36.4
10 9.1 45.5
12 10.9 54.6
Vu=256kips
10'-0

(1) Determine factored shear force. Use ACI Eq. (9-4) for wind loads only
Vu
= 1.6(160) = 256 kips
(2) Determine φVc
and maximum allowable φVn
From Table 6-5 φVc
= 7.3 ϫ 10 = 73 kips
φVn
= 36.4 ϫ 10 = 364 kips
Wall cross section is adequate (Vu
maximum φVn
); however, shear reinforcement is determined from
ACI Eq. (11-29) must be provided (Vu
φVc
).
(3) Determine required horizontal shear reinforcement
φVs
= Vu
- φVc
= 256 – 73 = 183 kips
φVs
= 183/10 = 18.3 kips/ft length of wall
Select horizontal bars from Table 6-4
For No. 5 @ 7 in., φVs
= 19.1 kips/ft 18.3 kips/ft O.K.
smax
= 18 in. 7 in. O.K.
Use No. 5 @ 7 in. horizontal reinforcement
Note: The use of minimum shear reinforcement No. 4 @ 10 in (Table 6-3) for an 8 in. wall thickness
is not adequate:
φVs
= 8.6 kips/ft only (Table 6-4).
(4) Determine required vertical shear reinforcement
ρ˜ = 0.0025 + 0.5(2.5 - hw/˜w)(ρt – 0.0025)
= 0.0025 + 0.5(2.5 – 2)(0.0055 – 0.0025)
= 0.0033
where hw/˜w = 20/10 = 2
ρt = Avh/s2h = 0.31/(8 ϫ 7) = 0.0055
Required Avn/s1 = ρ˜h = 0.0033 ϫ 8 = 0.026 in.2
/in.
For No. 5 bars: s1
= 0.31/0.026 = 11.9 in. 18 in. O.K.
Use No. 5 @ 12 in. vertical reinforcement.
6-8

6.4.2 Example 2: Design for Shear
For Alternate (2) of Building #2 (5-story flat plate), select shear reinforcement for the two shearwalls. Assume
that the total wind forces are resisted by the walls, with slab-column framing resisting gravity loads only.
(1) E-W direction
Total shear force at base of building (see Chapter 2, Section 2.2.2.1):
V = 6.9 + 13.4 + 12.9 + 12.2 + 12.6 = 58 kips
For each shearwall, V = 58/2 = 29 kips
Factored shear force (use ACI Eq. (9-4) for wind load only):
Vu
= 1.6(29) = 46.4 kips
For the E-W direction, assume that the shear force is resisted by the two 8 ft flange segments only. For
each segment:
φVc
= 7.3 ϫ 8 = 58.4 kips (see Table 6-5)
Since Vu for each 8 ft segment = 46.4/2 = 23.2 kips which is less than φVc/2 = 58.4/2 = 29.2 kips, provide
minimum wall reinforcement from Table 6-2 For 8 in. wall, use No. 4 @ 12 in. horizontal reinforcement
and No. 3 @ 11 in. vertical reinforcement.
(2) N-S direction
Total shear force at base of building (see Chapter 2, Section 2.2.2.1):
V = 16.2 + 31.6 + 30.6 + 29.2 + 30.7 = 138.3 kips
For each shearwall, V = 138.3/2 = 69.2 kips
6-9
8'-0
20'-8
8 typ.
Vu = 46.4 kips

Factored shear force:
Vu
= 1.6(69.2) = 110.6 kips
For the N-S direction, assume that the shear force is resisted by the 20 ft-8 in. web segment only. From
Table 6-5
φVc
= 7.3 ϫ 20.67 = 150.9 kips
Since φVc
/2 = 75.5 kips Vu
= 110.6 kips φVc
= 150.9 kips, provide minimum shear reinforcement from
Table 6-3 For 8 in. wall, use No. 4 @ 10 in. horizontal as well as vertical reinforcement.
(3) Check shear strength in 2nd story in the N-S direction
Vu
= 1.6(16.2 + 31.6 + 30.6 + 29.2)/2 = 86.1 kips
The minimum shear reinforcement given in Table 6-3 s still required in the 2nd story since φVc
/2 = 75.5
kips Vu
= 86.1 kips φVc
= 150.9 kips. For the 3rd story and above, the minimum wall reinforcement
given in Table 6-2 can be used for all wall segments (Vu
@ 3rd story = 62.7 kips φVc
/2 = 75.5 kips). For
horizontal reinforcement, use No. 4 @ 12 in., and for vertical reinforcement, use No. 3 @ 11 in.
(4) Summary of Reinforcement
Vertical bars: Use No. 4 @ 10 in. for 1st and 2nd stories*
No. 3 @ 10 in. for 3rd through 5th stories**
Horizontal bars: Use No. 4 @ 10 in. for 1st and 2nd stories
No. 4 @ 12 in. for 3rd through 5th stories
* Formoment strength, No. 6 @ 10 in. are required in the 8 ft. wall segments within the first story (see Example 6.5.1).
** Spacing of vertical bars reduced from 11 in. to 10 in. so that the bars in the 3rdstory can be splicedwith the bars in the 2ndstory.
6-10
8'-0
20'-8
8 typ.
Vu=110.6kips

6.5 DESIGN FOR FLEXURE
For buildings of moderate height, walls with uniform cross-sections and uniformly distributed vertical and
horizontal reinforcement are usually the most economical. Concentration of reinforcement at the extreme ends of
a wall (or wall segment) is usually not required except in high and moderate seismic zones. Uniform distribution
of the vertical wall reinforcement required for shear will usually provide adequate moment strength as well.
Minimum amounts of reinforcement will usually be sufficient for both shear and moment requirements.
In general, walls that are subjected to axial load or combined flexure and axial load need to be designed
as compression members according to the provisions given in ACI Chapter 10 (also see Chapter 5)*. For
rectangular shearwalls containing uniformly distributed vertical reinforcement and subjected to an axial load
smaller than that producing balanced failure, the following approximate equation can be used to determine the
nominal moment capacity of the wall6.1 (see Fig. 6-4):
where Ast
= total area of vertical reinforcement, in.2
˜w
= horizontal length of wall, in.
Pu
= factored axial compressive load, kips
fy
= yield strength of reinforcement = 60 ksi
Figure 6-4 Plan View of Shearwall for Approximate Nominal Moment Capacity
* In particular, ACI 10.2, 10.10, and 10.12 are applicable for walls.
φMn = φ 0.5Ast fy w 1+
Pu
Ast
fy
⎛
⎝
⎜
⎞
⎠
⎟ 1−
c
w
⎛
⎝
⎜
⎞
⎠
⎟
⎡
⎣
⎢
⎢
⎤
⎦
⎥
⎥
6-11
Ast
lw

h = thickness of wall, in.
φ = 0.90 (strength primarily controlled by flexure with low axial load)
Note that this equation should apply in a majority of cases since the wall axial loads are usually small.
6.5.1 Example: Design for Flexure
For Alternate (2) of Building #2 (5-story flat plate), determine the required amount of moment reinforcement
for the two shearwalls. Assume that the 8 ft wall segments resist the wind moments in the E-W direction and
the 20 ft-8 in. wall segments resist the wind moments in the N-S direction.
Roof: DL = 122 psf
Floors: DL = 142 psf
(1) Factored loads and load combinations
When evaluating moment strength, the load combination given in ACI Eq. (9-6) will govern.
U = 0.9D + 1.6W
(a) Dead load at first floor level:
Tributary floor area = 12 ϫ 40 = 480 sq ft/story
Wall dead load = (0.150 ϫ 3392)/144 = 3.53 kips/ft of wall height (see Sect. 6.3.1)
Pu = 0.9[(0.122 ϫ 480) + (0.142 ϫ 480 ϫ 4) + (3.53 ϫ 63)] = 498 kips
Proportion total Pu between wall segments:
2-8 ft segments: 2 ϫ 96 = 192 in. 192/440 = 0.44
1-20 ft-8 in. segment: 248 in. 248/440 = 0.56
For 2-8 ft segments: Pu = 0.44(498) = 219 kips
1-20 ft-8 in. segment Pu = 0.56(498) = 279 kips
(b) Wind moments at first floor level:
c
w
=
ω + α
2ω + 0.85β1
, where β1
= 0.85 for ʹfc
= 4000psi
ω =
Ast
w
h
⎛
⎝
⎜
⎞
⎠
⎟
fy
ʹfc
α =
Pu
w
h ʹfc
6-12
96
248

From wind load analysis (see Chapter 2, Section 2.2.2.1):
E-W direction
Mu
= 1.6 [(6.9 ϫ 63) + (13.4 ϫ 51) + (12.9 ϫ 39) + (12.2 ϫ 27) + (12.6 ϫ 15)]/2
= 1712 ft-kips/shearwall
N-S direction:
Mu
= 1.6 [(16.2 ϫ 63) + (31.6 ϫ 51) + (30.6 ϫ 39) + (29.2 ϫ 27) + (30.7 ϫ 15)]/2
(c) Values of Pu
and Mu
for the 2nd and 3rd floor levels are obtained in a similar manner:
For 2nd floor level: 2-8 ft segments: Pu
= 171 kips
1-20 ft-8 in. segment: Pu
= 218 kips
E-W direction: Mu
N-S direction: Mu
For 3rd floor level: 2-8 ft segment: Pu
= 128 kips
1-20 ft-8 in. segment: Pu
= 162 kips
E-W direction: Mu
N-S direction: Mu
(2) Design for Flexure in E-W direction
Initially check moment strength based on the required vertical shear reinforcement No. 4 @ 10 in.
(see Example 6.4.2).
(a) For 2-8 ft wall segments at first floor level:
Pu = 219 kips
Mu = 1712 ft-kips
˜w = 96 in.
For No. 4 @ 10 in. (2 wall segments):
6-13
Ast = 3.84 in.2
16
w = 96
Ast
= 2 × 0.24 × 8 = 3.84in.2
ω =
Ast
w h
⎛
⎝
⎜
⎞
⎠
⎟
fy
ʹfc
=
3.84
96 ×16
⎛
⎝
⎜
⎞
⎠
⎟
60
4
= 0.038
α =
Pu
w
h ʹfc
=
219
96 ×16 × 4
= 0.036

No. 4 @ 10 in. is not adequate for moment strength in the E-W direction at the first story level.
Try No. 5 @ 10 in.:
(b) For 2-8 ft wall segments at 2nd floor level:
Pu
= 171 kips
Mu
= 1016 ft-kips
Check No. 4 @ 10 in.:
No. 4 @ 10 in. (required shear reinforcement) is adequate for moment strength above the first floor.
(c) For 2-8 ft wall segments at 3rd floor level:
Pu
= 128 kips
Mu
= 580 ft-kips
Ast
= 3.84in.2
ω = 0.038
α =
171
96 × 16 × 4
= 0.028
c
w
=
0.038 + 0.028
2 0.038( ) + 0.72
= 0.082
Mn
= 0.5 × 3.84 × 60 × 96 1 +
171
3.84 × 60
⎛
⎝
⎜
⎞
⎠
⎟ 1 − 0.082( ) / 12 = 1474ft − kips
φMn
= 0.9 1474( ) = 1327ft − kips Mu
= 1016ft − kips O.K.
Ast
= 2 × 0.39 × 8 = 6.2in.2
ω =
6.2
96 ×16
⎛
⎝
⎜
⎞
⎠
⎟
60
4
= 0.061
c
w
=
0.061+ 0.036
2 0.061( )+ 0.72
= 0.114
Mn
= 0.5 × 6.2 × 60 × 96 1+
219
6.2 × 60
⎛
⎝
⎜
⎞
⎠
⎟ 1− 0.014( )/12 = 2094ft − kips
φMn
= 0.9 2094( ) = 1885ft − kips Mu
c
w
=
ω + α
2ω + 0.85 × 0.85( )
=
0.038 + 0.036
2 0.038( )+ 0.72
= 0.092
Mn
= 0.5Ast
fy w
1+
Pu
Ast
fy
⎛
⎝
⎜
⎞
⎠
⎟ 1−
c
w
⎛
⎝
⎜
⎞
⎠
⎟
= 0.5 × 3.84 × 60 × 96 1+
219
3.84 × 60
⎛
⎝
⎜
⎞
⎠
⎟ 1− 0.093 /12 = 1633ft − kips( )
φMn
= 0.9 1633( ) = 1469ft − kips Mu
= 1712ft − kips N.G.
6-14

Check No. 3 @ 10 in. (required shear reinforcement above 2nd floor):
No. 3 @ 10 in. (required shear reinforcement) is adequate for moment strength above the 2nd floor.
(3) Design for flexure in N-S direction
Initially check moment strength for required vertical shear reinforcement No. 4 @ 10 in.
(see Example 6.4.2)
(a) For 1-20 ft-8 in. wall segment at first floor level:
Pu
= 279 kips
Mu
= 4060 ft-kips
˜w
= 248 in.
h = 8 in.
For No. 4 @ 10 in.:
6-15
Ast
= 2 × 0.13× 8 = 2.08in.2
ω =
2.08
96 ×16
⎛
⎝
⎜
⎞
⎠
⎟
60
4
0.020
α =
128
96 ×16 × 4
= 0.021
c
w
=
0.020 + 0.021
2 0.020( )+ 0.72
= 0.054
Mn
= 0.5 × 2.08 × 60 × 96 1+
128
2.08 × 60
⎛
⎝
⎜
⎞
⎠
⎟ 1− 0.054( )/12 = 957ft − kips
φMn
= 0.9 957( ) = 861ft − kips Mu
Ast
= 0.24 × 20.67 = 4.96 in.2
ω =
4.96
248 × 8
⎛
⎝
⎜
⎞
⎠
⎟
60
4
= 0.038
α =
2.79
248 × 8 × 4
= 0.035
c
w
=
0.038 + 0.035
2 0.038( )+ 0.72
= 0.091
Mn
= 0.5 × 4.96 × 60 × 248 1+
279
4.96 × 60
⎛
⎝
⎜
⎞
⎠
⎟ 1− 0.091( )/12 = 5415 ft − kips
φMn
= 0.9 5415( ) = 4874 ft − kips Mu
= 4060 ft − kips O.K.
8
Ast = 4.96 in.2
w=248

(b) For one-20 ft-8 in. wall segment at 3rd floor level:
Pu
= 162 kips
Mu
= 1367 ft-kips
Check No. 3 @ 10 in. (required shear reinforcement above 2nd floor):
The required shear reinforcement for the 20 ft-8 in. wall segments is adequate for moment strength for full
height of building.
(4) Summary
Required shear reinforcement determined in Example 6.4.2 can be used for the flexural reinforcement
except for the 8 ft wall segments within the 1st floor where No. 5 @ 10 in. are required (see Fig. 6-5).
For comparison purposes, the shearwall was investigated using the program spColumn.6.3
For the reinforcement
shown in Fig. 6-5 at the 1st story level, the shearwall was analyzed for the combined factored axial load (due
to the dead loads) and moments (due to the wind loads) about each principal axis. The results are shown for
the x and y axes in Figs. 6-6 and 6-7, respectively. As expected, the load combination point (represented by
point 1 in the figures) is in the lower region of the interaction diagram, with the applied axial load well below
the balanced point. Since spColumn uses the entire cross-section when computing the moment capacity (and
not only certain segments as was done in the steps above), the results based on the reinforcement from the
approximate analysis will be conservative.
6-16
Ast
= 0.13× 20.67 = 2.96 in.2
ω =
4.96
248 × 8
⎛
⎝
⎜
⎞
⎠
⎟
60
4
= 0.020
α =
162
248 × 8 × 4
= 0.020
c
w
=
0.020 + 0.020
2 0.020( )+ 0.72
= 0.053
Mn
= 0.5 × 2.69 × 60 × 2.48 1+
162
2.96 × 60
⎛
⎝
⎜
⎞
⎠
⎟ 1− 0.053( )/12 = 3163ft − kips
φMn
= 0.9 3163( ) = 2847 ft − kips Mu
= 1367 ft − kips O.K.

6-17
#3 @ 10 in. (3rd thru 5th stories)
#4 @ 10 in. (2nd story)
#5 @ 10 in. (1st story)
#4 @ 10 in. (1st and 2nd story)
#4 @ 12 in. (3rd thru 5 stories)
8'-0
8typ.
20'-8
#4@10in.(1stand2ndstory)
#3@10in.(3rdthru5stories)
Figure 6-5 Required Reinforcement for Shearwall in Building #2

6-18
Figure 6-6 Interaction Diagram for Shearwall Bending About the X-axis
P (kip)
(Pmax)
7000
5000
3000
1000
-1000
10000 20000 30000 40000
Mx (k-ft)
(Pmin)
1

6-19
Figure 6-7 Interaction Diagram for Shearwall Bending About the Y-axis
P (kip)
(Pmax)
7000
5000
3000
1000
-1000
-10000 -6000 -2000 2000 6000 10000
My (k-ft)
(Pmin)
1
(Pmax)
(Pmin)

References
6.1 Cardenas, A.E., and Magura, D.D., “Strength of High-Rise Shear Walls—Rectangular Cross Section”,
SP-36, American Concrete Institute, SP-36, pp. 119-150.
6.2 Cardenas, A.E., Hanson, J.M., Corley, W.G., Hognestad, E., “Design Provisions for Shearwalls”, Journal
of the American Concrete Institute, Vol. 70, No. 3, March 1973, pp. 221-230.
6.3 spColumn, Design and investigation of reinforced concrete column sections, Structure Point, Skokie,
Illinois, 2010.
6-20

7-1
Chapter 7
Simplified Design for Footings
7.1 INTRODUCTION
A simplified method for design of spread footings is presented that can be used to obtain required footing
thickness with a one-step design equation based on minimum footing reinforcement. Also included are simplified
methods for shear, footing dowels, and horizontal load transfer at the base of a column. A simplified one-step
thickness design equation for plain concrete footings is also given. The discussion will be limited to the use of
individual square footings supporting square (or circular) columns and subject to uniform soil pressure. The design
methods presented are intended to address the usual design conditions for footings of low-to-moderate height
buildings. Footings that are subjected to uplift or overturning are beyond the scope of the simplified method.
Concrete specified strength for footing design must satisfy strength design requirements and durability
requirements (ACI Chapter 4). The design examples in this chapter assume concrete specified strength of
4000 psi. In cases where the foundation is exposed to freezing and thawing, deicing chemicals or severe levels
of sulfates, the required concrete strength should be determined based on ACI Chapter 4 requirements.
The simplified design equation provided in this chapter is applicable to a wide range of concrete strength.
Also, charts are provided for concrete strength ranging from 3 to 6 ksi.
7.2 PLAIN CONCRETE VERSUS REINFORCED CONCRETE FOOTINGS
Reinforced concrete footings are often used in smaller buildings without considering plain footings. Many
factors need to be considered when comparing the two alternatives, the most important being economic
considerations. Among the other factors are soil type, job-site conditions, and building size (loads to be
transferred). The choice between using reinforcement or not involves a trade-off between the amounts of
concrete and steel. The current market prices of concrete and reinforcement are important decision-making
parameters. If plain footings can save considerable construction time, then the cost of the extra concrete may
be justified. Also, local building codes should be consulted to determine if plain concrete footings are allowed
in certain situations. For a given project, both plain and reinforced footings can be quickly proportioned by the
simplified methods in this chapters and an overall cost comparison made (including both material and
construction costs). For the same loading conditions, the thickness of a plain footing will be about twice that
of a reinforced footing with minimum reinforcement (see Section 7.8).
7.3 SOIL PRESSURE
Soil pressures are usually obtained from a geotechnical engineer or set by local building codes. In cities where
experience and tests have established the allowable (safe) bearing pressures of various soils, local building codes

7-2
may be consulted to determine the bearing capacities to be used in design. In the absence of such information or
for conditions where the nature of the soils is unknown, borings or load tests should be made. For larger buildings,
borings or load tests should always be made.
In general, the base area of the footing is determined using unfactored loads and allowable soil pressures
(ACI 15.2.2), while the footing thickness and reinforcement are obtained using factored loads (ACI 15.2.1).
7.4 SURCHARGE
In cases where the top of the footing is appreciably below grade (for example, below the frost line) allowances
need to be made for the weight of soil on top of the footing. In general, an allowance of 100 pcf is adequate
for soil surcharge; unless wet packed conditions exist that warrant a higher value (say 130 pcf). Total surcharge
(or overburden) above base of footing can include the loads from a slab on grade, the soil surcharge, and the
footing weight.
7.5 ONE-STEP THICKNESS DESIGN FOR REINFORCED CONCRETE FOOTINGS
A simplified footing thickness equation can be derived for individual footings with minimum reinforcement
using the strength design data developed in Reference 7.1. The following derivation is valid for › = 4000 psi,
fy = 60,000 psi, and a minimum reinforcement ratio of 0.0018 (ACI 10.5.4).
For a 1ft wide design strip:
where Mu is in ft-kips.
Referring to Fig. 7-1, the factored moment Mu at the face of the column (or wall) is (ACI 15.4.2):
where c is the largest footing projection from face of column (or wall). Substituting Mu into the equation for
d2
reqd results in the following:
Mu = qu
c2
2
⎛
⎝
⎜
⎞
⎠
⎟ =
Pu
Af
c2
2
⎛
⎝
⎜
⎞
⎠
⎟
dreqd
2
=
Mu
φRn
=
Mu
×1000
0.9 ×117.9
= 9.43Mu
Set ρ = 0.0018 × 1.11 = 0.002*
Rn
= ρfy
1 −
0.5pfy
0.85 ʹfc
⎛
⎝
⎜
⎞
⎠
⎟
= 0.002 × 60,000 1 −
0.5 × 0.002 × 60,000
0.85 × 4000
⎛
⎝
⎜
⎞
⎠
⎟
= 117.9 psi
* The minimum value of ρ is multiplied by 1.11 to account for the ratio of effective depth d to overall thickness h, assumed as d/h ഡ 0.9.

7-3
Chapter 7 • Simplified Design for Footings
c
h
qu=Pu/Af
d
Figure 7-1 Reinforced Footing
c1 + d
Critical section, bo,
for two-way shear
bo = 4(c1 + d)
c1
c1d
c
Critical section for
beam shear
c1 + dbw
Tributary area for two-way shear
Tributary area for wide-beam shear
Figure 7-2 Tributary Areas and Critical Sections for Shear in Footings

* For square interior columns, Eq. (11-32) will govern where d/c1 ≤ 0.25. .
7-4
The above equation is in mixed units: Pu is in kips, c is in feet, Af is in square feet, and d is in inches.
The one-step thickness equation derived above is applicable for both square and rectangular footings (using largest
value of c) and wall footings. Since fy has a larger influence on d than does › , the simplified equation can be used
for other concrete strengths without a substantial loss in accuracy. As shown in Fig. 7-1, this derivation assumes
uniform soil pressure at the bottom of the footing; for footings subject to axial load plus moment, an equivalent
uniform soil pressure can be used.
According to ACI 11.11, the shear strength of footings in the vicinity of the column must be checked for both
one-way (wide-beam) shear and two-way shear. Fig. 7-2 illustrates the tributary areas and critical sections for
a square column supported by a square footing.
For wide-beam shear:
for normalweight concrete
where bw is the width of the footing, and Vu is the factored shear on the critical section (at a distance d from
the face of the column). In general,
The minimum depth d can be obtained from the following equation:
where qu
in psi
This equation is shown graphically in Fig. 7-3 for › = 3000, 4000, 5000 and 6000 psi.
For a footing supporting a square column, the two-way shear strength will be the lesser of the values of Vc
obtained fromACI Eqs. (11-32) and (11-33). Eq. (11-32) will rarely govern since the aspect ratio bo/d will usually be
considerably less than the limiting value to reduce the shear strength below .* Therefore, for two-way
shear,
where, for a square column, the perimeter of the critical section is bo = 4(c1 + d). The factored shear Vu on the
critical section (at d/2 from the face of the column) can be expressed as:
Vu
≤ 4φ ʹfc
bo
d
4 ʹfc
bo
d
d
c
=
qu
qu
+ 2φ ʹfc
Vu
= qu
bw
c − d( )
Vu
≤ 2φ ʹfc
bw
d
dreqd
2
= 4.7qu
c2
= 4.7
Pu
c2
Af
dreqd
= 2.2 qu
c2
= 2.2c
Pu
Af

7-5
where Ac = area of the column = c1
2
and vc = . Fig. 7-4 can be used to determine d for footings with
› = 3, 4, 5 and 6 ksi: given qu and Af/Ac, the minimum value of d/c1 can be read from the vertical axis.
Square footings, that are designed based on minimum flexural reinforcement will rarely encounter any one-
way or two-way shear problems when supporting square columns. For other footing and column shapes, shear
strength will more likely control the footing thickness. In any case, it is important to ensure that shear strength of
the footing is not exceeded.
7.5.1 Procedure for Simplified Footing Design
(1) Determine base area of footing Af from service loads (unfactored loads) and allowable (safe) soil pressure qa
determined for the site soil conditions and in accordance with the local building code.
4 ʹfc
Vu
= qu
Af
− c1
+ d( )
2
⎡
⎣
⎤
⎦
qu
4
+ φvc
⎛
⎝
⎜
⎞
⎠
⎟ d2
+
qu
2
+ φvc
⎛
⎝
⎜
⎞
⎠
⎟ c1
d −
qu
4
Af
− Ac( ) = 0

7-6
ʹ = 3ksi
fc
4ksi
5ksi
6ksi
Figure 7-3 Minimum d for Wide-Beam Shear

7-7
Figure 7-4(a) Minimum d for Two-Way Shear ( › = 3ksi)
30
45
60
75
90
105
120
135
150
165
180
195
210
225
240
= 15Af
Ac

7-8
Figure 7-4(b) Minimum d for Two-Way Shear ( › = 4ksi)
30
45
60
75
90
105
120
135
150
165
180
195
210
225
240
= 15Af
Ac

7-9
Figure 7-4(c) Minimum d for Two-Way Shear ( › = 5ksi)
30
45
60
75
90
105
120
135
150
165
180
195
210
225
240
= 15Af
Ac

7-10
Figure 7-4(d) Minimum d for Two-Way Shear ( › = 6ksi)
30
45
60
75
90
105
120
135
150
165
180
195
210
225
240
= 15Af
Ac

7-11
*
3 in. cover (ACI 7.7.1) + 1 bar diameter (ഡ1 in.) = 4 in.
**
For circular columns, c = distance from the face of an imaginary square column with the same area (ACI 15.3) to edge of footing.
The following equations for Pu usually govern (for live load ≤ 100 lb/ft2
):
Pu = 1.2D + 1.6L + 0.5Lr
ACI Eq. (9-2)
Pu = 1.2D + 1.6Lr
+ 0.5L ACI Eq. (9-3)
Pu = 1.2D ± 1.6W + 0.5L + 0.5Lr
ACI Eq. (9-4)
(2)Determine required footing thickness h from one-step thickness equation:
h = d + 4 in.*
where Pu = factored column load, kips
Af = base area of footing, sq ft
c = greatest distance from face of column to edge of footing, ft **
h = overall thickness of footing, in.
(3) Determine minimum d for wide-beam shear and two-way shear from Figs. 7-3 and 7-4, respectively. Use the
larger d obtained from the two figures, and compare it to the one obtained in step (2). In general, the value of d
determined in step (2) will govern. Note that it is permissible to treat circular columns as square columns with
the same cross-sectional area (ACI 15.3).
(4) Determine reinforcement:
As = 0.0018 bh
As per foot width of footing:
As = 0.022h (in.2
/ft)
Select bar size and spacing from Table 3-6. Note that the maximum bar spacing is 18 in. (ACI 7.6.5). Also, the
provisions in ACI 10.6.4, which cover the maximum bar spacing for crack control, do not apply to footings.
The size and spacing of the reinforcement must be chosen so that the bars can become fully developed. The
bars must extend at least a distance d from each face of the column, where d is the tension development length
of the bars (ACI 15.6). In every situation, the following conditions must be satisfied (see Fig. 7-5):
L 2 d + c1 + 6 in.
where L is the width of the footing and c1 is the width of the column.
Af
=
D + L + W + surcharge if any( )
qa
h = 2.2c
Pu
Af
+ 4in. ≥ 10in.

7-12
All of the spacing and cover criteria depicted in Fig. 7-6 are usually satisfied in typical situations; therefore,
d can be computed from the following (ACI 12.2):
For No. 6 and smaller:
For No. 7 and larger:
Where ψe
,ψe
and λ are factors depend on the bar location coating and concrete type (lightweight or normal-
weight). For normalweight concrete reinforced with uncoated bars located at the bottom of the footing, the
multiplier ψe
ψe
and the value of λ can be taken as 1.0.
Values of d for › = 3000 psi to 6000 psi are given in Table 7-1. In cases where the spacing and/or cover are
less than those given in Fig. 7-6, a more detailed analysis using the appropriate modification factors in ACI
12.2 must be performed to obtain d.
7.6 FOOTING DOWELS
7.6.1 Vertical Force Transfer at Base of Column
The following discussion addresses footing dowels designed to transfer compression forces only. Tensile
forces created by moments, uplift, or other causes must be transferred to the footings entirely by reinforcement
(ACI 15.8.1.2).
d
=
fy
ψt
ψe
20λ ʹfc
db
d
=
fy
ψt
ψe
25λ ʹfc
db
L
c13 3
Figure 7-5 Available Development Length for Footing Reinforcement

7-13
Compression forces must be transferred by bearing on concrete and by reinforcement (if required). Bearing
strength must be adequate for both column concrete and footing concrete. For the usual case of a footing with
a total area considerably larger than the column area, bearing on column concrete will always govern until
› of the column concrete exceeds twice that of the footing concrete (ACI 10.14.1). Design bearing strength of
concrete must not exceed φ(0.85›A1
), where A1
is the loaded area and φ = 0.65. For concrete strength
› = 4000 psi, the allowable bearing force Pnb
(in kips) on the column concrete is equal to Pnb
= 2.21 Ag
, where
Ag
is the gross area of the column in square inches. Values of Pnb
for concrete strengths 4, 5, and 6 ksi are listed
in Table 7-2 for the column sizes given in Chapter.
When the factored column load Pu exceeds the concrete bearing capacity φPnb the excess compression must be
transferred to the footing by reinforcement (extended column bars or dowels; see ACI 15.8.2). Total area of
reinforcement across the interface cannot be less than 0.5% of the column cross-sectional area (see Table 7-2).
For the case when dowel bars are used, it is recommended that at least 4 dowels (one in each corner of the
column) be provided.
#11 bars and smaller
(without epoxy coating)
db≥ 2.5db ≥5db
≥ 2db
Figure 7-6 Typical Spacing and Cover of Reinforcement in Footings
Table 7-1 Minimum Development Length d for Flexural Reinforcement in Footings (Grade 60)

7-14
Figure 7-7 shows the minimum dowel embedment lengths into the footing and column. For normalweight
concrete the dowels must extend into the footing a compression development length of , but
not less than 0.0003dbfy, where db is the diameter of the dowel bar (ACI 12.3.2)*. Table 7-3 gives the
minimum values of ˜dc for concrete with › = 3000, 4000, 5000 and 6000 psi**. The dowel bars are usually
extended down to the level of the flexural steel of the footing and hooked 90°as shown in Fig. 7-7. The hooks
are tied to the flexural steel to hold the dowels in place. It is important to note that the bent portions of
the dowels cannot be considered effective for developing the bars in compression. In general, the following
condition must be satisfied when hooked dowels are used:
h ˜dc + r + dbd + 2dbf + 3 in.
where r = minimum radius of dowel bar bend (ACI Table 7.2), in.
dbd = diameter of dowel, in.
dbf = diameter of flexural steel, in.
dc = 0.02db fy / ʹfc
Table 7-2 Bearing Capacity and Minimum Area of Reinforcement Across Interface
Table 7-3 Minimum Compression Development ˜dc Length for Grade 60 Bars
* The compression development length may be reduced by the applicable factor given in ACI 12.3.3.
** ˜dc can conservatively be taken as 22 db for all concrete with › 3000 psi.

7-15
For the straight dowels, the minimum footing thickness h must be ˜dc + 3 in.
In certain cases, the thickness of the footing must be increased in order to accommodate the dowels. If this is
not possible, a greater number of smaller dowels can be used.
For the usual case of dowel bars which are smaller in diameter than the column bars, the minimum dowel
embedment length into the column must be the larger of the compression development length of the column
bar (Table 7-3) or the compression lap splice length of the dowel bar (ACI 12.16.2). The splice length is
0.0005 fy db = 30 db for Grade 60 reinforcement, where db is the diameter of the dowel bar (ACI 12.16.1).
Table 7-4 gives the required splice length for the bar sizes listed. Note that the embedment length into the column
is 30 db when the dowels are the same size as the column bars.
#4 thru #11 column bars
#4 thru #11 dowels
h
Minimumdowel
embedmentinto
column
3 clear 3 clear + 2dbf + dbd + r
dbf = diameter of flexural steel
dbd = diameter of dowel
r = Minimum radius of dowel bar bend
dbd
dbf
Figure 7-7 Footing Dowels

7-16
Bar
Size
Lap Splice Length
(in.)
# 4 15
# 5 19
# 6 23
# 7 26
# 8 30
# 9 34
#10 38
#11 42
* '
f 3000 psic
Table 7-4 Minimum Compression Lap Splice Length for Grade 60 Bars*
7.6.2 Horizontal Force Transfer at Base of Column
ACI 11.6.7 permits permanent compression to function like shear friction reinforcement. For all practical cases in
ordinary buildings, the column dead load will be more than enough to resist any column shear at the top of the footing.
For normalweight concrete the horizontal force Vu to be transferred cannot exceed the smallest of φ(0.2f’c
Ac),
φ(480 + 0.08f’c
)Ac and φ 1600 Ac in pounds where Ac is gross area of column(ACI 11.6.5). For a column
concrete strength › = 4000 psi, this maximum force is equal to φ(800Ac). Dowels required to transfer
horizontal force must have full tensile anchorage into the footing and into the column (ACI 11.6.8). The values
given in Table 8-1 and 8-5 can be modified by the appropriate factors given in ACI 12.2.
7.7 EXAMPLE: REINFORCED CONCRETE FOOTING DESIGN
Design footings for the interior columns of Building No. 2 (5-story flat plate). Assume base of footings located
5 ft below ground level floor slab (see Fig. 7-8). Permissible soil pressure qa = 6 ksf.
Top of floor slab
at ground level
5'-0N.T.S.
Figure 7-8 Interior Footing for Building No. 2

7-17
(1) Design Data:
Service surcharge = 50 psf
Assume weight of soil and concrete above footing base = 130 pcf
Interior columns: 16 in. ϫ 16 in. (see Examples 5.7.1 and 5.7.2)
4-No.8 bars (non-sway frame)
8-No.10 bars (sway frame)
f’c = 4000 psi (column)
f’c = 4000 psi (footing)
(2) Load combinations
a) gravity loads (Alternate (2)): PDL
= 351 kips
PLL
= 56.4 kips
Mservice
= 6.6 ft-kips (10.6/1.6)
Pu
= 512 kips (ACI Eq. 9-2)
Mu
= 10.6 ft-kips
b) gravity loads + wind (Alternate (1)): PDL
= 339.4 kips
PLL
= 56.4 kips
Mservice
= 75.4 ft-kips
Pu
= 498 kips (ACI Eq. 9-3)
Mu
= 66 ft-kips
Or
Pu
= 436 kips (ACI Eq. 9-4)
Mu
= 113 ft-kips
(3) Base area of footing
Determine footing base area for gravity loads only, then check footing size for gravity plus winds loads.
Total weight of surcharge = (0.130 ϫ 5) + 0.05 = 0.70 ksf
Net permissible soil pressure = 6 – 0.70 = 5.3 ksf
Try 9 ft-6 in. ϫ 9 ft-6 in. square footing (Af = 90.25 sq ft)
Check gravity plus wind loading for 9 ft-6 in. ϫ 9 ft-6 in. footing
Af = 90.25 sq ft
Sf = bh2/6 = (9.5)3/6 = 142.9 ft3
q =
P
Af
+
M
Sf
=
339.4 + 56.4
90.25
+
75.4
142.9
= 4.9 5.3 O.K.
Af =
351+ 56.4
5.3
= 76.9 sq ft*
* Neglect small moment due to gravity loads.

7-18
(4) Footing thickness
Footing projection
c = [(9.5 – 16/12)]/2 = 4.08 ft
Try h = 27 in. (2 ft-3 in.)
Check if the footing thickness is adequate for shear:
d ഡ 27– 4 = 23 in.
For wide-beam shear, use Fig. 7-3. With qu
= = 5.7 ksf, read d/c ഡ 0.33. Therefore, the minimum d is
d = 0.33 ϫ 4.08 = 1.35 ft = 16.2 in. 23 in. O.K.
Use Fig. 7-4 for two-way shear:
Interpolating between Af/Ac = 45 and 60, read d/c1 1.13 for qu = 5.7 ksf. The minimum d for two-way shear is:
d = 1.13 ϫ 16 = 18.1 in. 23 in. O.K.
Therefore, the 27 in. footing depth (d = 23 in.) is adequate for flexure and shear.
(5) Footing reinforcement
As = 0.022 h = 0.022(27) = 0.59 in.2
/ft
Try No.7 @ 12 in. (As = 0.60 in.2/ft; see Table 3-7)
Determine the development length of the No.7 bars (see Fig. 7-6):
cover = 3 in. 2db = 2 ϫ 0.875 = 1.8 in.
side cover = 3 in. 2.5 ϫ 0.875 = 2.2 in.
clear spacing = 12 – 0.875 = 11.1 in. 5 x 0.875 = 4.4 in.4.4 in.
Since all of the cover and spacing criteria given in Fig. 7-6 are satisfied, Table 7-1 can be used to deter-
mine the minimum development length.
For › = 4000 psi: d = 42 in.
Check available development length:
L = 9.5 ϫ 12 = 114 in. (2 ϫ 42) + 16 + 6 = 106 in. O.K.
Use 9 ft-6 in. ϫ 9 ft-6 in. square footing (L=118 in.)
Af
Ac
=
90.25
162
/144( )
= 50.8
h = 2.2c
Pu
Af
+ 4 in. = 2.2 4.08( )
512
90.25
+ 4 = 25.4 in. 10 in. O.K.
512
90.25

7-19
Total bars required:
Use 10-No.7, 9 ft-4 in. long (each way)*
(6) Footing dowels
Footing dowel requirements are different for sway and non-sway frames. For the sway frame, with wind
moment transferred to the base of the column, all of the tensile forces produced by the moment must be
transferred to the footing by dowels. The number and size of dowel bars will depend on the tension development
length of the hooked end of the dowel and the thickness of the footing. The dowel bars must also
be fully developed for tension in the column.
For the non-sway frame, subjected to gravity loads only, dowel requirements are determined as follows:**
(a) For 16 ϫ 16 in. column (Table 7-2):
φPnb = 566 kips
Minimum dowel area = 1.28 in.2
Since φPnb Pu = 512 kips, bearing on concrete alone is adequate for transfer of compressive force.
Use 4-No.6 dowels (As = 1.76 in.2
)
(b) Embedment into footing (Table 7-3):
For straight dowel bars,
h ≥ db + 3 in.
db = 17 in. for No.6 dowels with f’c = 3000 psi
h = 27 in. 17 + 3 = 20 in. O.K.
For hooked dowel bars,
h ≥ db+ r + dbd + 2dbf + 3 in.
r = 3dbd = 3 ϫ 0.75 = 2.25 in. (ACI Table 7.2)
h = 27 in. 17 + 2.25 + 0.75 + (2 ϫ 0.875) + 3 = 24.75 in. O.K.
122 − 6
12
= 9.66 spaces
* 13-No.6 or 8-No.8 would also be adequate.
** The horizontal forces produced by the gravity loads in the first story columns are negligible; thus, dowels required for
vertical load transfer will be adequate for horizontal load transfer as well.

7-20
(c) Embedment into column:
The minimum dowel embedment length into the column must be the larger of the following:
- compression development length of No.8 column bars (› = 4000 psi) = 19 in. (Table 7-3)
- compression lap splice length of No.6 dowel bars = 23 in. (Table 7-4) (governs)
For No.6 hooked dowels, the total length of the dowels is
23 + [27 – 3 – (2 ϫ 0.875)] = 45.25 in.
Use 4-No.6 dowels ϫ 3 ft-10 in.
Figure 7-9 shows the reinforcement details for the footing in the non-sway frame
7.8 ONE-STEP THICKNESS DESIGN FOR PLAIN CONCRETE FOOTINGS
Depending on the magnitude of the loads and the soil conditions, plain concrete footings may be an econom-
ical alternative to reinforced concrete footings. Structural plain concrete members are designed according to ACI
318-11, Chapter 22.7.2
. For normalweight plain concrete, the maximum moment design strength is
(ACI Eq. 22-2).With φ = 0.55 (ACI 318, Section 9.3.5)
A simplified one-step thickness design equation can be derived as follows (see Fig. 7-10):
For a one-foot design strip:
qu
c2
c
⎛
⎝
⎜
⎞
⎠
⎟ ≤ 5φ ʹfc
h2
6
⎛
⎝
⎜
⎞
⎠
⎟
hreqd
2
= 3qu
c2
5φ ʹfc
=
0.6Pu c2
Af
φ ʹfc
Mu
≤ 5φ ʹfc
Sm
5φ ʹfc
Sm
2'-0
4-#6 dowels x 3'-10
(90° standard end hook)
10-#7 bars x 9'-4
(each way)
3 clear
3 clear
4-#8 column bars
2'-3
9'-10 square
Figure 7-9 Reinforcement Details for Interior Column in Building No.2 (Non-sway Frame)

7-21
To allow for unevenness of excavation and for some contamination of the concrete adjacent to the soil, an
additional 2 in. in overall thickness is required for plain concrete footings; thus,
The above footing thickness equations are in mixed units:
Pu = factored column load, kips
Af = base area of footing, sq ft
c = greatest distance from face of column to edge of footing, ft (ACI 318, Section 22.7.5)
h = overall thickness of footing, in. 8 in. (ACI 318, Section 22.7.4)
Thickness of plain concrete footings will be controlled by flexural strength rather than shear strength for the
usual proportions of plain concrete footings. Shear rarely will control. For those cases where shear may cause
concern, the nominal shear strength is given in ACI 22.5.4.
7.8.1 Example: Plain Concrete Footing Design
For the interior columns of Building No.2, (Braced Frame), design a plain concrete footing.
From Example 7.7:
Af = 9 ft-10 in. ϫ 9 ft-10 in. = 96.7 sq ft
Pu = 512 kips
c = footing projection = 4.25 ft
For ʹfc
= 3000 psi: h = 4.5c
Pu
Af
+ 2in.
For ʹfc
= 4000 psi: h = 4.15c
Pu
Af
+ 2in.
Pu
c
h
qu = Pu/Af
Figure 7-10 Plain Concrete Footing

7-22
For f’c = 4000 psi:
Bearing on column:
Allowable bearing load = 0.85φf’cAg, φ = 0.55 ACI 318, Section 22.5.5
= 0.85 ϫ 0.55 ϫ 4 ϫ 162
= 479 kips Pu 501 kips
The excess compression must be transferred to the footing by reinforcement consisting of extended column
bars or dowels (calculations not show).
Figure 7-11 illustrates the footing for this case.
References
7.1 Notes on ACI 318-08, Chapter 7, Design for Flexure, 10th Edition, EB708, Portland Cement Association,
Skokie, Illinois, 2008.
7.2 Building Code Requirements for Structural Concrete ACI 318-11 and Commentary, American Concrete
Institute, Farmington Hills, Michigan, 2011.
h = 4.15c
Pu
Af
+ 2in. = 4.15 4.25( )
512
96.7
+ 2 = 40.6 + 2 = 42.6in.
4-#8 column bars
2'-0
4-#6 dowels x 4'-0
3'-9
9'-6 square
Figure 7-11 Plain Concrete Footing for Interior Column of Building No. 2 (Non-sway Frame)

8-1
Chapter 8
Structural Detailing of
Reinforcement for Economy
8.1 INTRODUCTION
Structurally sound details and proper bar arrangements are vital to the satisfactory performance of reinforced
concrete structures. The details and bar arrangements should be practical, buildable, and cost-effective.
Ideally, the economics of reinforced concrete should be viewed in the broad perspective, considering all facets
in the execution of a project. While it may be important to strive for savings in materials, many engineers often
tend to focus too much on material savings rather than on designing for construction efficiencies. No doubt,
savings in material quantities should result from a highly refined “custom design” for each structural member
in a building. However, such a savings in materials might be false economy if significantly higher construction
costs are incurred in building the custom-designed members.
Trade-offs should be considered in order to minimize the total cost of construction, including the total in-place
cost of reinforcement. Savings in reinforcement weight can be traded-off for savings in fabrication, placing,
and inspection for overall economy.
8.2 DESIGN CONSIDERATIONS FOR REINFORCEMENT ECONOMY
The following notes on reinforcement selection and placement will usually provide for overall economy and
may minimize costly project delays and job stoppages:
(1) First and foremost, show clear and complete reinforcement details and bar arrangements in the
Contract Documents. This issue is addressed in Section 1.1 of ACI Detailing Manual8.1: “…the
responsibility of the Engineer is to furnish a clear statement of design requirements; the responsibility
of the [Reinforcing Steel] Detailer is to carry out these requirements.” ACI 318 further emphasizes that
the designer is responsible for the size and location of all reinforcement and the types, locations, and
lengths of splices of reinforcement (ACI 1.2.1 and 12.14.1).
(2) Use Grade 60 reinforcing bars. Grade 60 bars are the most widely used and are readily available in all
sizes up to and including No.11; No.14 and No.18 bars are not generally inventoried in regular stock.
Also, bar sizes smaller than No.6 generally cost more per pound and require more placing labor per
pound of reinforcement.
(3) Use straight bars only in flexural members. Straight bars are regarded as standard in the industry. Truss
(bent) bars are undesirable from a fabrication and placing standpoint and structurally unsound where
stress reversals occur.

(4) In beams, specify bars in single layers only. Use one bar size for reinforcement on one face at a given
span location. In slabs, space reinforcement in whole inches, but not at less than 6-in. spacing.
(5) Use largest bar sizes possible for the longitudinal reinforcement in columns. Use of larger bars sizes
and fewer bars in other structural members will be restricted by code requirements for development of
reinforcement, limits on maximum spacing, and distribution of flexural reinforcement.
(6) Use or specify fewest possible bar sizes for a project.
(7) Stirrups are typically the smaller bar sizes, which usually result in the highest total in-place cost of
reinforcement per ton. For overall economy and to minimize congestion of reinforcement, specify the
largest stirrup bar size (fewest number of stirrups) and the fewest variations in spacing. Stirrups spaced
at the maximum allowable spacing, are usually the most economical.
(8) When closed stirrups are required for structural integrity, and torsion does not govern the design,
specify two-piece closed types (conforming to ACI 12.13.5) to facilitate placing unless closed stirrups
are required for torsion.
(9) Fit and clearance of reinforcing bars warrant special attention by the Engineer. At beam-column joints,
arrangement of column bars must provide enough space or spaces to permit passage of beam bars.
Bar details should be properly prepared and reconciled before the bars are fabricated and delivered to
the job site. Member joints are far too important to require indiscriminate adjustments in the field to
facilitate bar placing.
(10) Use or specify standard reinforcing bar details and practices:
• Standard end hooks (ACI 7.1). Note that the tension development length provisions in ACI 12.5 are
only applicable for standard hooks conforming to ACI 7.1.
• Typical bar bends (see ACI 7.2 and Fig. 6 in Ref. 8.1).
• Standard fabricating tolerances (Fig. 8 in Ref. 8.1). More restrictive tolerances must be indicated by
the Engineer in the Contract Documents.
• Tolerances for placing reinforcing bars (ACI 7.5). More restrictive tolerances must be indicated by
the Engineer in the Contract Documents.
Care must be exercised in specifying more restrictive tolerances for fabricating and placing reinforcing
bars. More restrictive fabricating tolerances are limited by the capabilities of shop fabrication
equipment. Fabricating and placing tolerances must be coordinated. Tolerances for the formwork must
also be considered and coordinated.
(11) Never permit field welding of crossing reinforcing bars for assembly of reinforcement (“tack” welding,
“spot” welding, etc.). Tie wire will do the job without harm to the bars.
8-2

(12) Avoid manual arc-welded splices of reinforcing bars in the field wherever possible, particularly for
smaller projects.
(13) A frequently occurring construction problem is having to make field corrections to reinforcing bars
partially embedded in hardened concrete. Such “job stoppers” usually result from errors in placing or
fabrication, accidental bending caused by construction equipment, or a design change. Field bending
of bars partially embedded in concrete is not permitted except if such bending is shown on the design
drawings or authorized by the Engineer (ACI 7.3.2). ACI R7.3 offers guidance on this subject. Further
guidance on bending and straightening of reinforcing bars is given in Reference 8.2.
8.3 REINFORCING BARS
Billet-steel reinforcing bars conforming to ASTM A 615, Grade 60, are the most widely used type and grade
in the United States. Combining the Strength Design Method with Grade 60 bars results in maximum overall
economy. This design practice has made Grade 60 reinforcing bars the standard grade. The current edition of
ASTM A 615 reflects this practice, as only bar sizes No.3 through No.6 in Grade 40 are included in the
specification. Also listed are Grade 75 bars in sizes No.6 through No.18 only. The larger bar sizes in Grade 75
(No.11, No.14, and No.18) are usually used in columns made of high strength concrete in high-rise buildings.
The combination of high strength concrete and Grade 75 bars may result in smaller column sizes, and, thus,
more rentable space, especially in the lower levels of a building. It is important to note that Grade 75 bars may
not be readily available in all areas of the country; also, as mentioned above, No.14 and No.18 bars are not
commonly available in distributors’stock. ACI 3.5.3.2 permits the use of Grade 75 bars provided that they meet
all the requirements listed in that section (also see ACI 9.4).
When important or extensive welding is required, or when more bendability and controlled ductility are
required (as in seismic construction*), use of low-alloy reinforcing bars conforming to ASTM A 706 should be
considered. Note that the specification covers only Grade 60 bars. Local availability should be investigated
before specifying A 706 bars.
8.3.1 Coated Reinforcing Bars
Zinc-coated (galvanized) and epoxy-coated reinforcing bars are used increasingly for corrosion-protection in
reinforced concrete structures. An example of a structure that might use coated bars is a parking garage where
vehicles track in deicing salts.
Zinc-coated (galvanized) reinforcing bars must conform to ASTM A 767; also, the reinforcement to be coated
must conform to one of the specifications listed in ACI 3.5.3.1. Bars are usually fabricated before galvanizing.
In these cases, the minimum finished bend diameters given in Table 2 of ASTM A 767 must be specified.
ASTM A 767 has two classes of coating weights of surface. Class I (3.0 oz./Sq ft for No.3bars and 3.5 oz./sq
ft for No.4 and larger bars) is normally specified for general construction. ASTM A 767, requires that sheared
ends to be coated with a zinc-rich formulation. Also when bars are fabricated after galvanizing, ASTM A 767
8-3
Chapter 8 • Structural Detailing of Reinforcement for Economy
* ACI 21.1.5.2 specifically requires reinforcing bars complying with ASTM A 706 to be used in frame members and in wall
boundary elements subjected to seismic forces. Note that ASTM A 615 Grade 40 and Grade 60 bars are also allowed if they
meet all of the requirements in the section.

requires that the damaged coating be repaired with a zinc-rich formulation. If ASTM A 615 billet-steel bars
are being supplied, ASTM A 767 requires that a silicon analysis of each heat of steel be provided. It is
recommended that the above pertaining requirements be specified when fabrication after galvanization
includes cutting and bending.
Uncoated reinforcing steel (or any other embedded metal dissimilar to zinc) should not be permitted in the
same concrete element with galvanized bars, nor in close proximity to galvanized bars, except as part of a
cathodic-protection system. Galvanized bars should not be coupled to uncoated bars.
Epoxy-coated reinforcing bars must conform to ASTM A 775, and the reinforcement to be coated must
conform to one of the specifications listed in ACI 3.5.3.1. The film thickness of the coating after curing shall
be 7 to 12 mils (0.18 to 0.30 mm). Also, there shall not be more than an average of one holidays (pinholes not
discernible to the unaided eye) per linear foot of the coated bar.
Proper use of ASTM A 767 and A 775 requires the inclusion of provisions in the project specifications for the
following items:
• Compatible tie wire, bar supports, support bars, and spreader bars in walls.
• Repair of damaged coating after completion of welding (splices) or installation of mechanical
connections.
• Repair of damaged coating after completion of field corrections, when field bending of coated bars
partially embedded in concrete is permitted.
• Minimizing damage to coated bars during handling, shipment, and placing operations; also, limits
on permissible coating damage and, when required, repair of damaged coating.
Reference 8.3 contains suggested provisions for preceding items for epoxy-coated reinforcing bars.
8.4 DEVELOPMENT OF REINFORCING BARS
8.4.1 Introduction
The fundamental requirement for development (or anchorage) of reinforcing bars is that a reinforcing bar
must be embedded in concrete a sufficient distance on each side of a critical section to develop the peak
tension or compression stress in the bar at the section. The development length concept in ACI 318 is based on
the attainable average bond stress over the length of embedment of the reinforcement. Standard end hooks
or mechanical devices may also be used for anchorage of reinforcing bars, except that hooks are effective for
developing bars in tension only (ACI 12.1.1).
8.4.2 Development of Deformed Bars in Tension
The ACI provides two equations for development length calculations:
8-4

where
˜d = development length, in.
db = nominal diameter of bar, in.
fy = specified yield strength for bar, psi
› = specified compressive strength of concrete, psi
ψt = reinforcement location factor
= 1.3 for horizontal reinforcement so placed that more than 12 in. of fresh concrete is cast below
the bar being developed or spliced
= 1.0 for other reinforcement
ψe = coating factor
= 1.5 for epoxy-coated bars with cover less than 3db or clear spacing less than 6db
= 1.2 for all other epoxy-coated bars
= 1.0 for uncoated reinforcement
The product of ψt and ψe need not be taken greater than 1.7.
λ = lightweight aggregate concrete factor
= 1.3 when lightweight aggregate concrete is used, or
= 1.0 for normal weight concrete
The above equations are valid for clear spacing of bars being developed or spliced not less than db,
clear cover not less than db and stirrups or ties not less than the code minimum, or clear spacing not
less than 2db and clear cover not less than db (ACI 12.2.2). For cases where the reinforced bars are
closely spaced, or the provided cover is less than db, the development length must be increased ACI
R12.2.
For 4000 psi normal weight concrete and uncoated reinforcing bottom bars:
˜d = 38 db for No. 6 and smaller bars
˜d = 48 db for No. 7 and larger bars
d
=
fy
ψ t
ψ e
25λ ʹfc
⎛
⎝
⎜
⎞
⎠
⎟ db
For No. 6 and smaller bars
d
=
fy
ψ t
ψ e
20λ ʹfc
⎛
⎝
⎜
⎞
⎠
⎟ db
For No. 7 and smaller bars
8-5

The development length may be reduced when the provision for excess reinforcement given in ACI 12.2.5 are
satisfied. Also ˜d must never be taken less than 12 in. (ACI 12.2.1).
Values for tension development length ˜d are given in Table 8-1 for grade 60 reinforcing bars and concrete
strengths ranging from 3 to 6 ksi. The values in the table are based on bars that are not epoxy-coated and on
normal weight concrete. To obtain ˜d for top bars (horizontal bars with more than 12 in. of concrete cast below
the bars) the tabulated values must be multiplied by 1.3 (ACI 1212.2.4). The cover and clear spacing referred
to in the table are depicted in Fig. 8-1.
As can be seen from the table, very long development lengths are required for the larger bar sizes, especially
when the cover is less than db.
8-6
cover, c
side
cover
Transverse
reinforcement
Clear space
s
db
Figure 8-1 Cover and Clear Spacing of the Reinforcement
Table 8-1 Tension Development Length for ˜d Grade 60 Bars
Values are based on bars which are not epoxy-coated and on normal weight concrete. For top
Bars, multiply tabulated values by 1.3

8.4.3 Development of Hooked Bars in Tension
The ACI provides the following equation for development length in tension for bars ending with standard hook:
The values for ψe and λ are as defined before.
For normal weight concrete with › = 4000 psi and uncoated reinforcing bars with fy = 60,000 psi the development
length ˜dh = 19 db. Table 8-2 lists the development length ˜dh for different bar sizes and concrete strengths.
Table 8-2 Minimum Development Lengths ˜dh for Grade 60 Bars with Standard End Hooks (in.)*
dh =
0.02fy
ψ e
λ ʹfc
⎛
⎝
⎜
⎞
⎠
⎟ db
8-7
Standard 90° hook
2 min.
cover
Asfy
db
12db
˜dh
Standard 180° hook
4db or
2.5 min.
db
4db
5db
˜dh
Asfy
#3 through #8
#9, #10, #11
*Values based on normal weight concrete.

The development lengths give in Table 8-2 are applicable for end hooks with side cover normal to plane of
hook of not less than 21
⁄2 in. and end cover (90° hooks only) of not less than 2 in. For these cases, ˜dh = 0.7˜hb,
but not less than 8db or 6 in. (ACI 12.5.1). For hooked bar anchorage in beam-column joints, the hooked beam
bars are usually placed inside the vertical column bars, with side cover greater than the 21
⁄2-in. minimum
required for application of the 0.7 reduction factor. Also, for 90° end hooks with hook extension located inside
the column ties, the 2-in. minimum end cover will usually be satisfied to permit the 0.7 reduction factor.
Where development for full fy is not specifically required, the tabulated values of ˜dh in Table 8-2 may be
further reduced for excess reinforcement (ACI 12.5.3.d). As noted above, ˜dh must not be less than 8db or 6 in.
ACI 12.5.4 provides additional requirements for hooked bars terminating at the discontinuous end of members
(ends of simply supported beams, free end of cantilevers, and ends of members framing into a joint where the
member does not extend beyond the joint). If the full strength of the hooked bar must be developed, and if both
the side cover and the top (or bottom) cover over the hook is less than 21
⁄2 in., closed ties ore stirrups spaced at
3db maximum are required along the full development length ˜dh. The reduction factor in ACI 12.5.3.b must not
be used in this case. At discontinuous ends of slabs with confinement provided by the slab continuous on both
sides normal to the plane of the hook, the requirements in ACI 12.5.4 for confining ties or stirrups do not apply.
8.4.4 Development of Bars in Compression
Shorter development lengths are required for bars in compression than in tension since the weakening effect
of flexural tension cracks in the concrete is not present. The development length for deformed bars in
compression is , but not less than 0.0003dbfy or 8 in. (ACI 12.3). For concrete with
› = 4000 psi and grade 60 reinforcement bars ˜dc = 19db. The minimum development for bars in compression
is 8 in. Table 8-3 lists the development length in compression for grade 60 bars and concrete strengths ranging
from 3000 to 6000 psi. The values may be reduced by the applicable factors in ACI 12.3.3.
8.5 SPLICES OF REINFORCING BARS
Three methods are used for splicing reinforcing bars: 1) lap splices, 2) welded splices, and 3) mechanical
connections. The lap splice is usually the most economical splice. When lap splices cause congestion or
dc = 0.02db fy / ʹfc
8-8
Table 8-3 Minimum Compression Development Length ˜dc for Grade 60 Bars (in.)

field placing problems, mechanical connections or welded splices should be considered. The location of
construction joints, provision for future construction, and the particular method of construction may also make
lap splices impractical. In columns, lapped offset bars may need to be located inside the column above to
reduce reinforcement congestion; this can reduce the moment capacity of the column section at the lapped
splice location because of the reduction in the effective depth. When the amount of vertical reinforcement is
greater than 4%, and/or when large factored moments are present, use of butt splices—either mechanical
connections or welded splices—should be considered in order to reduce congestion and to provide for greater
nominal moment strength of the column section at the splice location.
Bars in flexural members may be spliced by non-contact lap splices (ACI 12.14.2.3); however, contact lap
splices are preferred since the bars are tied and are less likely to displace when the concrete is placed.
Welded splices generally require the most expensive field labor. For projects of all sizes, manual arc-welded
splices will usually be the most costly method of splicing due to the costs of inspection.
Mechanical connections are made with proprietary splice devices. Performance information and test data should
be obtained directly from the manufacturers. Basic information about mechanical connections and the types of
proprietary splice devices currently is available from Reference 8.4. Practical information on splicing and
recommendations for the design and detailing of splices are given in Reference 8.5.
8.5.1 Tension Lap Splices
Tension lap splices are classified as Class A or Class B (ACI 12.15.1). The minimum lap length for a Class A
splice is 1.0˜d, and for a Class B splice it is 1.3˜d, where ˜d is the tension development length of the bars. When
calculating the development length ˜d the factor in ACI 12.2.5 for excess reinforcement must not be used, since
the splice classifications already reflect any excess reinforcement at the splice location. Also the minimum of
12 in. for ˜d does not apply for lap splice.
The minimum lap lengths for Class A splices can be obtained from Table 8-1. For Class B splices, the
minimum lap lengths are determined by multiplying the values from Table 8-1 by 1.3. The effective clear
spacing between splices bars is illustrated in Fig. 8-2. For staggered splices in slabs or walls, the effective clear
spacing is the distance between adjacent spliced bars less the diameters of any intermediate unspliced bars
(Fig. 8-2a), The clear spacing to be used for splices in columns with offset bars and for beam bar splices are
shown in Figs. 8-2b and 8-2c, respectively.
In general, tension lap splices must be Class B except that Class A splices are allowed when both of the
following conditions are met: 1) the area of reinforcement provided is at least twice that required by analysis
over the entire length of the splice and 2) one-half or less of the total reinforcement is spliced within the
required lap length (ACI 12.15.2). Essentially, Class A splices may be used at locations where the tensile stress
is small. It is very important to specify which class of tension splice is to be used, and to show clear and
complete details of the splice in the Contract Documents.
8.5.2 Compression Lap Splices
Minimum lengths for compression lap splices is 0.0005fydb for fy of 60,000 psi or less (ACI 12.16.1).
Minimum lap splice length for Grade 60 bars in normal weight concrete are given in Table 8-4. The values
8-9

apply for all concrete strengths greater than or equal to 3000 psi. For Grade 60 bars, the minimum lap splice
length is 30 db but not less than 12 in. When bars of different size are lap spliced, the splice length shall be the
larger of 1) development length of larger bar, or 2) splice length of smaller bar (ACI 12.16.2). For columns, the
lap splice lengths may be reduced by a factor of 0.83 when the splice is enclosed throughout its length by ties
specified in ACI 12.17.2.4. The 12 in. minimum lap length also applies.
8-10
Table 8-4 Minimum Compression Lap
Splice Lengths for Grade 60 Bars*
Bar
size
Minimum lap length
(in.)
# 3 12
# 4 15
# 5 19
# 6 23
# 7 26
# 8 30
# 9 34
#10 38
#11 42
3000 psi› ≥
lap length
(a) Wall and slab reinforcement
db
db
s1
s1
s = (s1 – db)
s = (s1 – db)
(b) Beam bar splices(b) Column with offset corner bars
Spliced bars
Spliced bars
s
s
s
s
Figure 8-2 Effective Clear Spacing of Spliced Bars
*

8.6 DEVELOPMENT OF FLEXURAL REINFORCEMENT
8.6.1 Introduction
The requirements for development of flexural reinforcement are given in ACI 12.10, 12.11, and 12.12. These
sections include provisions for:
• Bar extensions beyond points where reinforcement is no longer required to resist flexure.
• Termination of flexural reinforcement in tension zones.
• Minimum amount and length of embedment of positive moment reinforcement into supports.
• Limits on bar sizes for positive moment reinforcement at simple supports and at points of inflection.
• Amount and length of embedment of negative moment reinforcement beyond points of inflection.
Many of the specific requirements are interdependent, resulting in increased design time when the provisions
are considered separately. To save design time and costs, recommended bar details should be used. As was
discussed earlier in this chapter, there is potential overall savings in fabrication, placing, and inspection costs
when recommended bar details are used.
8.6.2 Requirements for Structural Integrity
The preceding chapters covered structure design under code prescribed conventional loads. Structures may
suffer local damage from severe local abnormal loads, or events which are not considered in the design. Such
loads or events include explosions due to gas or industrial liquids, vehicle impact; impact of falling objects,
and local effects of very high winds such as tornadoes. The overall integrity of a reinforced concrete structure
to withstand such abnormal loads can be substantially enhanced by providing relatively minor changes in the
detailing of the reinforcement. ACI 7.13 provides detailing provisions intended to improve the redundancy and
ductility of structures. This is achieved by providing, as a minimum, some continuity reinforcement or tie
between horizontal framing members. In the event of damage to a major supporting element or an abnormal
loading event, the integrity reinforcement is intended to confine any resulting damage to a relatively small area,
thus improving overall stability. The provisions of ACI 7.13 are not an alternate to design for major abnormal
loads and events such as; blast load and progressive collapse. Design for these abnormal loads is beyond the
scope of this publication.
8.6.3 Recommended Bar Details
Recommended bar details (including provisions of ACI 7.13) for continuous beams, one-way slabs, one-way
joist construction, and two-way slabs (without beams) are given in Figs. 8-3 through 8-6. Similar details can
be found in References 8.1 and 8.6. The figures may be used to obtain bar lengths for members subjected to
uniformly distributed gravity loads only; adequate bar lengths must be determined by analysis for members
subjected to lateral loads. Additionally, Figs. 8-3 through 8-5 are valid for beams, one-way slabs, and
one-way joists that may be designed by the approximate method given in ACI 8.3.3.* Fig. 8-6 can be used to
determine the bar lengths for two-way slabs without beams.**
8-11
* Undernormal conditions, the barlengths give in Figs. 8-3 through 8-5 will be satisfactory. However, forspecial conditions, amore
detailedanalysis will be required. In any situation, it is the responsibility of the engineerto ensure that adequate barlengths are provided.
** To reduce placing andinspection time, all of the top bars in the column strip of atwo-way slab system can have the same length at
aparticularlocation (either0.30 ˜n forflat plates or0.33 ˜n forflat slabs), insteadof the two different lengths shown in Figs. 8-6(a)
and8-6(c).

8-12
One piece closed stirrups or U-stirrups
with 135∞ hook required for shear type
Std. hook
(typ.)
Note 2
Note 1
00
00
6
0
0
0.125ñ3
ñ3ñ2ñ1
0.125ñ20.125ñ1 0.125ñ2
(a) Beams without closed stirrups
(b) Perimeter beams
Notes: (1) Larger of 1/4(A+
or mechanical or welded splice (ACI 7.13.2.2 and 7.13.2.3)
(2) Larger of 1/6(A-
s1) or 1/6(A-
s2) but not less than two bars continuous or spliced with Class B splices
or mechanical or welded splice (ACI 7.13.2.2)
s1) or 1/4(A+
s2) but not less than two bars continuous or spliced with Class B splices
s2A+
s1A+
s3A+
s2A- s3A-
s1A-
≥1/4(A+
s1)
Larger of 3 or 3
ñ2 ñ3
Larger of 3 or 3
ñ1 ñ2
ñ1
4
Std. hook
(typ.)
Larger of 3 or 3
ñ1 ñ2
Larger of 3 or 3
ñ2 ñ3
s3A+
s2A+
s1A+
0.125ñ1 0.125ñ2 0.125ñ30.125ñ2
ñ3ñ2ñ1
Note 1 0 00
ñ1
4
≥1/4(A+
s1)
6
Std. hook
or anchored
to develop
Fy
Figure 8-3 Recommended Bar Details for Beams
8.7 SPECIAL BAR DETAILS AT SLAB-TO-COLUMN CONNECTIONS
When two-way slabs are supported directly by columns (as in flat plates and flat slabs), transfer of moment
between slab and column takes place by a combination of flexure and eccentricity of shear (see Chapter 4,
Section 4.4.1). The portion of the unbalanced moment transferred by flexure is assumed to be transferred over
a width of slab equal to the column width c plus 1.5 times the slab thickness h on either side of the column.
For edge and interior columns, the effective slab width is (c+3h), and for corner columns it is (c+1.5h). An
adequate amount of negative slab reinforcement is required in this effective slab width to resist the portion of
the unbalanced moment transferred by flexure (ACI 13.5.4). In some cases, additional reinforcement must be
concentrated over the column to increase the nominal moment resistance of the section. Note that minimum
bar spacing requirements must be satisfied at all locations in the slab (ACI 13.3.2). Based on recommendations
in Reference 8.7, examples of typical details at edge and corner columns are shown in Figs. 8-7 and 8-8.

8-13
Temperature
shrinkage
reinforcement
Larger of 0.30ñ1 or 0.30ñ2 Larger of 0.30ñ2 or 0.30ñ3
Std.
hook 0.25ñ1
0.125ñ1 0.125ñ2 0.125ñ2 0.125ñ3
ñ3ñ2ñ1
6666
6
0 00
≥1/4(A+
s1) ≥1/4(A+
s2)
≥1/4(A+
s3)A+
s1 A+
s2 A+
s3
Note: At least one bar continuous or spliced with a Class B splice or mechanical or welded splice
(ACI 7.13.2.1)
0.125ñ3
ñ3
0
0.125ñ2
ñ2
0.125ñ2
0.125ñ1
ñ1
006
0.25ñ1 Slab
reinforcement
Note
Distribution rib
(if required)
(a) Joist bar details
2 for 20 forms
2.5 for 30 forms
Face of support
3'-01.5Form depth
Top slab
Slab
reinforcement
Joist width
(b) Joist section (c) Plan of standard tapered
end where required
Slab
reinforcement
Form depth
Top slab
#4 cont.
4 min.
(e) Distribution rib (if required)(d) Plan of standard square end joists
Single
bar with
standard
hook or
anchored
to develop
Fy
Figure 8-5 Recommended Bar Details for One-Way Joist Construction
Figure 8-4 Recommended Bar Details for One-Way Slabs

8-14
Note 1
Note 2
Note 3
Note 4 Note 4
Std.
hook
(typ.)
(a) Column strip
0
00
6
0
ñ3ñ2ñ1
s3A+
s2A-
s3A-
s1A-
≥ 0.5(A+
s1)
s1A+ s2A+
6
≥ 0.5(A+
s2)
0.30ñ1
0.20ñ1
˜1 ˜2 ˜3
6
≥ 0.5( s2A- )≥ 0.5( s1A- )
Notes: (1) Larger of 0.30ñ1 or 0.30ñ2
(2) Larger of 0.20ñ1 or 0.20ñ2
(3) At least two bars continuous or anchored per ACI 13.3.8.5
(4) Splices are permitted in this region
Continuous
bars
Std.
hook
(typ.)
00
6
0
ñ3ñ2ñ1
s3A+
˜1 ˜2 ˜3
0.15˜20.15˜20.15˜1 0.15˜3
0.22ñ1
(b) Middle strip
Figure 8-6 Recommended Bar Details for Two-Way Slabs (Without Beams)

8-15
(8 + 3) #4
Columnstrip
c+3h
0.3˜n
s/2
s/2
s/2ssssss
Start1stbaroneachside@
fromedgeofcolumnstrip.
Spaceremainder@s.
Design Drawing
Start 1st bar @ column centerline (if uniformly spaced bar is on centerline,
start additional bars @ 3 on each side). Provide 3 min. spacing from
uniformly spaced bars (if possible).
Notes: (1) Maximum spacings s = 2 x slab thickness ≤ 18 in. (ACI 13.3.2)
(2) Where additional top bars are required, show the total number of bars
on the design drawing as (8 + 3) #4 where 8 indicates the number of
uniformly spaced bars and 3 indicates the number of additional bars.
Figure 8-7 Example of a Typical Detail for Top Bars at Edge Columns (Flat Plate)
Start 1st additional bar @ 6 from edge. Space
remainder @ 3 (if possible). Provide 3 min. spacing
from uniformly spaced bars (if possible).
Design Drawing
(5 + 3) #4
Columnstrip
c+1.5h
Start 1st bar @ 3 from edge.
Space remainder @ s.
ssss
0.3˜n
6
3
Figure 8-8 Example of a Typical Detail for Top Bars at Corner Columns (Flat Plate)

8.8 SPECIAL SPLICE REQUIREMENTS FOR COLUMNS
8.8.1 Construction and Placing Considerations
For columns in multistory buildings, one-story high preassembled reinforcement cages are usually used. It is
common practice to locate the splices for the vertical column bars just above the floor level. In certain
situations, it may be advantageous to use two-story high cages since this will reduce the number of splices and,
for lap splices, will reduce the amount of reinforcing steel. However, it is important to note that two-story high
cages are difficult to brace; the required guy wires or projecting bars may interfere with other construction
operations such as the movement of cranes for transporting equipment and material. Also, it is more difficult
and time-consuming to place the beam or girder bars at the intermediate floor level since they have to be
threaded through the column steel. These two reasons alone are usually more than sufficient to offset any
expected savings in steel that can be obtained by using two-story high cages. Thus, one-story high cages are
usually preferred.
8.8.2 Design Considerations
Special provisions for column splices are given in ACI 12.17. In general, column splices must satisfy
requirements for all load combinations for the column. For example, column design will frequently be
governed by the gravity load combination (all bars in compression). However, the load combination, which
includes wind loads, may produce tensile stresses in some of the bars. In this situation, a tension splice is
required even though the load combination governing the column design did not produce any tensile stresses.
When the bar stress due to factored loads is compressive, lap splices, butt-welded splices, mechanical
connections, and end-bearing splices are permitted. Table 8-4 may be used to determine the minimum
compression lap splice lengths for Grade 60 bars. Note that these lap splice lengths may be multiplied by 0.83
for columns with the minimum effective area of ties (throughout the splice length) given in ACI 12.17.2.4.
In no case shall the lap splice length be less than 12 in. Welded splices and mechanical connectors must meet
the requirements of ACI 12.14.3. A full welded splice, which is designed to develop in tension, at least
1.25 Abfy (Ab = area of bar) will be adequate for compression as well. A full mechanical connection must
develop in compression (or tension) at least 1.25 Abfy. End-bearing splices transfer the compressive stresses by
bearing of square cut ends of the bars held in concentric contact by a suitable device (ACI 12.16.4). These types
of splices may be used provided the splices are staggered or additional bars are provided at splice locations (see
ACI 12.17.4 and the following discussion).
A minimum tensile strength is required for all compression splices. A compression lap splice with a length
greater than or equal to the minimum value given in ACI 12.16.1 has a tensile strength of at least 0.25 Abfy.
As noted above, full welded splices and full mechanical connectors develop at least 1.25 Abfy in tension. For
end-bearing splices, the continuing bars on each face of the column must have a tensile strength of 0.25 Asfy
where As is the total area of steel on the face of the column. This implies that not more than three-quarters of
the bars can be spliced on each face of the column at any one location. Consequently, to ensure minimum
tensile strength, end-bearing splices must be staggered or additional bars must be added if more than three-
quarters of the bars are to be spliced at any one location.
8-16

8-17
Lap splices, welded splices, and mechanical connections are permitted when the bar stress is tensile;
end-bearing splices must not be used (ACI 12.16.4.1). According to ACI 12.14.3, full welded splices and full
mechanical connections must develop in tension at least 1.25 Abfy. When the bar stress on the tension face of
the column is less than or equal to 0.5 fy, lap splices must be Class B if more than one-half of the bars are
spliced at any section, or Class A if half or fewer of the bars are spliced and alternate splices are staggered by
the tension development length ˜d (ACI 12.17.2). Class B splices must be used when the bar stress is greater
than 0.5 fy.
Lap splice requirements for columns are illustrated in Fig. 8-9. For factored load combinations in Zone 1, all
column bars are in compression. In Zone 2, the bar stress fs on the tension face of the column varies from zero
to 0.5 fy in tension. For load combinations in Zone 3, fs is greater that 0.5 fy. The load combination that
produces the greatest tensile stress in the bars will determine which type of lap splice is to be used.
Load-moment design charts (such as the one in Figs. 5-17 through 5-24 in Chapter 5) can greatly facilitate the
design of lap splices for columns.
Typical lap splice details for tied columns are shown in Fig. 8-10. Also given in the figure are the tie spacing
requirements of ACI 7.8 and 7.10.5 (see Chapter 5). When a column face is offset 3 in. or more, offset bent
longitudinal bars are not permitted (ACI 7.8.1.5). Instead, separate dowels, lap spliced with the longitudinal
bars adjacent to the offset column faces must be provided. Typical splice details for footing dowels are given
in Chapter 7, Fig. 7-7.
Column Framing with Structural Walls (Non-sway Frame)
In this example, the required lap splice length will be determined for an interior column in the 2nd story; the
splice will be located just above the 9 in. floor slab at the 1st level.
(1) Column Size and Reinforcement
In Example 5.7.2, a 16 ϫ 16 in. column size was established for the entire column stack. It was
determined that 4-No.8 bars were required in both the 1st and 2nd floor columns.
(2) Lap Splice Length
Since the columns carry only gravity loads, all of the column bars will be in compression (Zone 1 in
Fig. 8-9). Therefore, a compression lap splice is sufficient.
From Table 8-4, the minimum compression lap splice length required for the No.8 bars is 30 in. In this
situation, No.3 ties are required @ 16 in. spacing.
According to ACI 12.17.2.4, the lap splice length may be multiplied by 0.83 if ties are provided with
an effective area of 0.0015hs throughout the lap splice length.

8-18
All bars in compression
0 fs 0.5fy
on tension face
fs 0.5fy
on tension face
Moment
Zone 3
Load
Zone 1
fs = 0.5fy
in tension
fs = 0.5fy
compression
fs fy
fs = fy
fs = 0
Zone 2
Figure 8-9 Special Splice Requirements for Columns

8-19
Slab
Slab
A A
Lapsplice
s/2 (ACI 7.10.5.4)
s/2 (ACI 7.10.5.4)
Additional ties within 6 in. of offset
(ACI 7.8.1.3)
Tie spacing, s (ACI 7.10.5.2)
Offset bend
(ACI 7.8.1)
s/2 (ACI 7.10.5.4)
Beams on all
four sides of
column
3 max. (ACI 7.10.5.5)
Additional ties within 6 in. of offset
(ACI 7.8.1.3)
Offset bend
(ACI 7.8.1)
Slab
Section A-A
Figure 8-10 Column Splice and Tie Details

Two legs of the No.3 ties are effective in each direction; thus, the required spacing s can be determined
as follows:
2 ϫ 0.11 = 0.22 in.2 = 0.0015 ϫ 16 ϫ s
or, s = 9.2 in.
Splice length = 0.83 ϫ 30 = 24.9
Use a 2 ft-1 in. splice length with No.3 ties @ 9 in. throughout the splice length.
Column Framing Without Structural Walls (Sway Frame)
As in Example 8.8.3, the required lap splice length will be determined for an interior column in the 2nd
story, located just above the slab at the 1st level.
(1) Load Data
In Example 5.7.1, the following load combinations were obtained for the 2nd story interior columns:
Gravity loads: Pu = 399 kips ACI Eq. (9-2)
Mu = 13 ft-kips
Gravity + Wind loads: Pu = 399 kips ACI Eq. (9-3)
Mu = 47 ft-kips
Pu = 371 kips ACI Eq. (9-4)
Mu = 77 ft-kips
Pu = 243 kips ACI Eq. (9-6)
Mu = 68 ft-kips
(2) Column Size and Reinforcement
A 16 ϫ 16 in. column size was established in Example 5.7.1 for the 1st story columns. This size is used
for the entire column stack; the amount of reinforcement can be decreased in the upper levels. It was
also shown that the 1st story columns were slender, and the 8-No.10 bars were required for the
factored axial loads and magnified moments at this level.
The reinforcement for the 2nd story columns can be determined using the procedure outlined in
Chapter 5. As was the case for the 1st story columns, the 2nd story columns are slender (use k = 1.2;
see Chapter 5):
k u
r
=
1.2 12 ×12( )− 8.5⎡⎣ ⎤⎦
0.3×16
= 34 22
8-20

8-21
(3) Lap Splice Length
The load combination represented by point 8 (ACI Eq. (9-6)) in Fig. 8-11 governs the type of lap splice
to be used, since it is the combination that produces the greatest tensile stress fs in the bars. Note that
the load combination represented by point 6 (ACI Eq. (9-4)) which governed the design of the column
does not govern the design of the splice. Since 0 fs 0.5 fy at point 8, a Class B splice must be used
(all the bars spliced ACI 12.17.2.2).
Required splice length = 1.3 ˜d where ˜d is the tension development length of the No.10 bars (of the
lower column).
Clear bar spacing Х 3.5 in. = 2.8 db (see Fig. 8-2)
Cover db = 1.27 in.
From Table 8-1, ˜d = 60.2 in.
1.3˜d = 1.3 ϫ 60.2 = 78.3 in.
Thus, a 6 ft-6 in. splice length would be required which is more than one-half of the clear story height.
Decreasing the bar size in the 1st story columns would result in slightly smaller splice lengths; how-
ever, the reinforcement ratio would increase from 4% (8-No.10) to 4.7% (12-No.9). Also, labor costs
would increase since more bars would have to be placed and spliced.
One possible alternative would be to increase the column size. For example, a 18 ϫ 18 in. column
would require about 8-No.8 bars in the 1st story. It is important to note that changing the dimensions
of the columns would change the results from the lateral load analysis, affecting all subsequent calcu-
lations; a small change, however, should not significantly alter the results.
References
8.1 ACI Detailing Manual – 2004, SP-66(04), American Concrete Institute, Farmington Hills, Michigan, 2004,
(PCA LT185).
8.2 Stecich, J.P., Hanson, J.M., and Rice, P.F., “Bending and Straightening of Grade 60 Reinforcing Bars,”
Concrete International: Design Construction, VOl. 6, No. 8, August 1984, pp. 14-23.
8.3 “Suggested Project Specifications Provisions for Epoxy-Coated Reinforcing Bars,” Engineering Data
Report No. 19, Concrete Reinforcing Steel Institute, Schaumburg, Illinois, 1984.
8.4 Types of Mechanical Splices of Reinforcing Bars, ACI Committee 439, American Concrete Institute,
Farmington Hills, Michigan, 2007, 22 pp.

8.5 Reinforcement: Anchorages and Splices, 4th Edition, Concrete Reinforcing Steel Institute, Schaumburg,
Illinois, 1997, 100 pp.
8.6 CRSI Handbook, Concrete Reinforcing Steel Institute, Schaumburg, Illinois, 10th Edition, 2008.
8.7 “Design of Reinforcement for Two-Way Slab-to-Column Frames Laterally Braced or Not Braced,”
Structural Bulletin No. 9, Concrete Reinforcing Steel Institute, Schaumburg, Illinois, June 1993.
8-22

9-1
Chapter 9
Design Considerations for
Economical Formwork
9.1 INTRODUCTION
Depending on a number of factors, the cost of formwork can be as high as 60% of the total cost of a cast-in-
place concrete structure. For this reason, it is extremely important to devise a structural system that
will minimize the cost of formwork. Basic guidelines for achieving economical formwork are given in
Reference 9.1, and are summarized in this chapter.
Formwork economy should initially be considered at the conceptual stage or the preliminary design phase of
a construction project. This is the time when architectural, structural, mechanical, and electrical systems are
conceived. The architect and the engineer can help reduce the cost of formwork by following certain basic
principles while laying out the plans and selecting the structural framing for the building. Additional savings
can usually be achieved by consulting a contractor during the initial design phases of a project.
Design professionals, after having considered several alternative structural framing systems and having
determined those systems that best satisfy loading requirements as well as other design criteria, often make
their final selections on the concrete framing system that would have the least amount of concrete and possibly
the least amount of reinforcing steel. This approach can sometimes result in a costly design. Complex
structural frames and nonstandard member cross sections can complicate construction to the extent that any
cost savings to be realized from the economical use of in-place (permanent) materials can be significantly
offset by the higher costs of formwork. Consequently, when conducting cost evaluations of concrete structural
frames, it is essential that the costs of formwork be included.
9.2 BASIC PRINCIPLES TO ACHIEVE ECONOMICAL FORMWORK
There is always the opportunity to cut costs in any structural system. The high cost of formwork relative to the
costs of the other components makes it an obvious target for close examination. Three basic design principles
that govern formwork economy for all site-cast concrete structures are given below.
9.2.1 Standard Forms
Since most projects do not have the budget to accommodate custom forms, basing the design on readily
available standard form sizes is essential to achieve economical formwork. Also, designing for actual
dimensions of standard nominal lumber will significantly cut costs. A simplified approach to formwork
carpentry means less sawing, less piecing together, less waste, and less time; this results in reduced labor and
material costs and fewer opportunities for error by construction workers.

9.2.2 Repetition
Whenever possible, the sizes and shapes of the concrete members should be repeated in the structure. By doing
this, the forms can be reused from bay to bay and from floor to floor, resulting in maximum overall savings.
The relationship between cost and changes in depth of horizontal construction is a major design consideration.
By standardizing the size or, if that is not possible, by varying the width and not the depth of beams, most
requirements can be met at a lowered cost, since the forms can be reused for all floors. To accommodate load and
span variations, only the amount of reinforcement needs to be adjusted. Also, experience has shown that changing
the depth of the concrete joist system from floor to floor because of differences in superimposed loads actually
results in higher costs. Selecting different joist depths and beam sizes for each floor may result in minor savings in
materials, but specifying the same depth for all floors will achieve major savings in forming costs.
9.2.3 Simplicity
In general, there are countless variables that must be evaluated and then integrated into the design of a building.
Traditionally, economy has meant a time-consuming search for ways to cut back on quantity of materials.
As noted previously, this approach often creates additional costs—quite the opposite effect of that intended.
An important principle in formwork design is simplicity. In light of this principle, the following questions
should be considered in the preliminary design stage of any project:
(1) Will custom forms be cost-effective? Usually, when standard forms are used, both labor and materials costs
decrease. However, custom forms can be as cost-effective as standard forms if they are required in a
quantity that allows mass production.
(2) Are deep beams cost-effective? As a rule, changing the beam depth to accommodate a difference in load
will result in materials savings, but can add considerably to forming costs due to field crew disruptions and
increased potential for field error. Wide, flat beams are more cost-effective than deep narrow beams.
(3) Should beam and joist spacing be uniform or vary with load? Once again, a large number of different
spacings (closer together for heavy loads, farther apart for light) can result in material savings. However, the
disruption in work and the added labor costs required to form the variations may far exceed savings in materials.
(4) Should column size vary with height and loading? Consistency in column size usually results in reduced
labor costs, particularly in buildings of moderate height. Under some rare conditions, however, changing
the column size will yield savings in materials that justify the increased labor costs required for forming.
(5) Are formed surface tolerances reasonable? Section 3.4 of ACI Standard 3479.2
provides a way of
quantitatively indicating tolerances for surface variations due to forming quality. The suggested tolerances
for formed cast-in-place surfaces are shown in Table 9-1 (Table 3.1 of ACI 347). The following simplified
guidelines for specifying the class of formed surface will usually minimize costs: a) Class A finish should
be specified for surfaces prominently exposed to public view, b) Class B finish should be specified for
surfaces less prominently exposed to public view, c) Class C finish should be specified for all noncritical
or unexposed surfaces, and d) Class D finish should be specified for concealed surfaces where roughness
is not objectionable. If a more stringent class of surface is specified than is necessary for a particular
formed surface, the increase in cost may become disproportionate to the increase in quality; this is
illustrated in Fig. 9-1.
9-2

Table 9-1 Permitted Irregularities in Formed Surfaces Checked with a 5-ft Template9.1
Figure 9-1 Class of Surface Versus Cost
9.3 ECONOMICAL ASPECTS OF HORIZONTAL FRAMING
Floors and the required forming are usually the largest cost component of a concrete building structure. The
first step towards achieving maximum economy is selecting the most economical floor system for a given plan
layout and a given set of loads. This will be discussed in more detail below. The second step is to define a
regular, orderly progression of systematic shoring and reshoring. Timing the removal of the forms and
requiring a minimum amount of reshoring are two factors that must be seriously considered since they can have
a significant impact on the final cost.
Figures 1-5 and 1-6 show the relative costs of various floor systems as a function of bay size and superimposed
load. Both figures are based on a concrete strength › = 4000 psi. For a given set of loads, the slab system that
is optimal for short spans is not necessarily optimal for longer spans. Also, for a given span, the slab system
that is optimal for lighter superimposed loads is not necessarily optimal for heavier loads. Reference 9.3
provides material and cost estimating data for various floor systems. It is also very important to consider the
fire resistance of the floor system in the preliminary design stage (see Chapter 10). Required fire resistance
ratings can dictate the type of floor system to specify in a particular situation.
The relationship between span length, floor system, and cost may indicate one or more systems to be
economical for a given project. If the system choices are equally cost-effective, then other considerations
(architectural, aesthetic, etc.) may become the determining factor.
Beyond selection of the most economical system for load and span conditions, there are general techniques that
facilitate the most economical use of the chosen system.
9.3.1 Slab Systems
Whenever possible, avoid offsets and irregularities that cause a “stop and start” disruption of labor and require
additional cutting (and waste) of materials (for example, breaks in soffit elevation). Depressions for terrazzo,
9-3
Chapter 9 • Design Considerations for Economical Formwork
Class of surface
A B C D
1/8 in. 1/4 in. 1/2 in. 1 in.
Quality
Class A
Class B
Class C
Cost

tile, etc. should be accomplished by adding concrete to the top surface of the slab rather than maintaining a
constant slab thickness and forming offsets in the bottom of the slab. Cross section (a) in Fig. 9-2 is less costly
to form than cross section (b).
Figure 9-2 Depressions in Slabs
When drop panels are used in two-way systems, the total depth of the drop h1
should be set equal to the actual
nominal lumber dimension plus 3
/4
-in. for plyform (see Fig. 9-3). Table 9-2 lists values for the depth h1
based
on common nominal lumber sizes. As noted above, designs which depart from standard lumber dimensions are
expensive.
Figure 9-3 Formwork for Drop Panels
Table 9-2 Drop Panel Depth, h1
9-4
(a)
h
(b)
hh
h1
Slab
Plyform Drop
16’-6” or greater
*Keep drop dimensions constant
Nominal lumber
size
Actual lumber
size (in.)
Plyform
thickness (in.)
h1
(in.)
2X 1½ ¾ 2¼
4X 3½ ¾ 4¼
6X 5½ ¾ 6¼
8X 7¼ ¾ 8

Whenever possible, a minimum 16 ft (plus 6 in. minimum clearance) spacing between drop panel edges should
be used (see Fig. 9-3). Again, this permits the use of 16 ft long standard lumber without costly cutting of
material. For maximum economy, the plan dimensions of the drop panel should remain constant throughout the
entire project.
9.3.2 Joist Systems
Whenever possible, the joist depth and the spacing between joists should be based on standard form
dimensions (see Table 9-3).
The joist width should conform to the values given in Table 9-3 also. Variations in width mean more time for
interrupted labor, more time for accurate measurement between ribs, and more opportunities for jobsite error;
all of these add to the overall cost.
Table 9-3 Standard Form Dimensions for One-Way Joist Construction (in.)
It is extremely cost-effective to specify a supporting beam with a depth equal to the depth of the joist. By doing
this, the bottom of the entire floor system can be formed in one horizontal plane. Additionally, installation costs
for utilities, partitions, and ceilings can all be reduced.
9.3.3 Beam-Supported Slab Systems
The most economical use of this relatively expensive system relies upon the principles of standardization and
repetition. Of primary importance is consistency in depth and of secondary importance is consistency in width.
These two concepts will mean a simplified design; less time spent interpreting plans and more time for field
crews to produce.
9-5
Width Depth Flange width Width of joist
20 8, 10, 12 7/8, 2 ½ 5, 6
30 8, 10, 12, 14, 16, 20 7/8, 3 5, 6, 7
53 16, 20 3½ 7, 8, 9, 10
66 14, 16, 20 3 6, 7, 8, 9, 10
Width of joist=2 x flange width*
*Applies to flange widths 7/8 in.
Width of joist
Width
Flange width
12
1
Depth

9.4 ECONOMICAL ASPECTS OF VERTICAL FRAMING
9.4.1 Walls
Walls provide an excellent opportunity to combine multiple functions in a single element; by doing this, a more
economical design is achieved. With creative layout and design, the same wall can be a fire enclosure for stair
or elevator shafts, a member for vertical support, and bracing for lateral loads. Walls with rectangular
cross-sections are less costly than nonrectangular walls.
9.4.2 Core Areas
Core areas for elevators, stairs, and utility shafts are required in many projects. In extreme cases, the core may
require more labor than the rest of the floor. Standardizing the size and location of floor openings within the
core will reduce costs. Repeating the core framing pattern on as many floors as possible will also help to
minimize the overall costs.
9.4.3 Columns
Although the greatest costs in the structural frame are in the floor system, the cost of column formwork should
not be overlooked. Whenever possible, use the same column dimensions for the entire height of the building.
Also, use a uniform symmetrical column pattern with all of the columns having the same orientation. Planning
along these general lines can yield maximum column economy as well as greater floor framing economy
because of the resulting uniformity in bay sizes.
9.5 GUIDELINES FOR MEMBER SIZING
9.5.1 Beams
• For a line of continuous beams, keep the beam size constant and vary the reinforcement from span to span.
• Wide flat beams (same depth as joists) are easier to form than beams projecting below the bottom of the
joints (see Fig, 9-4).
Figure 9-4 One-Way Joist Floor System
9-6
Beam
Joist
Wide flat beams are
more ecomonical than
narrow, deep beams

• Spandrel beams are more cost intensive than interior beams due to their location at the edge of a floor
slab or at a slab opening. Fig. 9-5 lists some of the various aspects to consider when designing these members.
Figure 9-5 Spandrel Beams
• Beams should be as wide as, or wider than, the columns into which they frame (see Fig. 9-6). In addition
for formwork economy, this also alleviates some of the reinforcement congestion at the intersection.
Figure 9-6 Beam-Column Intersections
• For heavy loading or long spans, a beam deeper than the joists may be required. In these situations, allow
for minimum tee and lugs at sides of beams as shown in Fig. 9-7. Try to keep difference in elevation
between bottom of beam and bottom of floor system in modular lumber dimensions.
9-7
Overhangs difficult
and costly
Upturned spandrel
may be more
economical
Spandrel projection
beyond face of column
difficult and costly
Narrow deep beam can cause
difficulty in placing concrete and/or
rebars; widen to minimize
problems
Greatest formwork economy
achieved when beam is the same
width as the column

Figure 9-7 One-Way Joist Floor System with Deep Beams
9.5.2 Columns
• For maximum economy, standardize column location and orientation in a uniform pattern in both
directions (see Fig, 9-8).
Figure 9-8 Standard Column Location and Orientation for a Typical Bay
• Columns should be kept the same size throughout the building. If size changes are necessary, they should
occur in 2 in. increments, one side at a time (for example, a 22 ϫ 22 in. column should go to a 24 x 22 in.,
then to a 24 ϫ 24 in., etc.) Gang forming can possibly be used when this approach to changing columns
sizes is utilized. When a flying form system is used, the distance between column faces and the flying
form must be held constant. Column size changes must be made parallel to the flying form.
9-8
Beam
Joist
Use actual lumber
dimensions (see
Table 9-2)
Keep beam width larger than
or equal to column
*Allow for lug on each side of deep beam
Constant
dimension
Vary this dimension
(only as necessary)
Typical bay

• Use the same shape as often as possible throughout the entire building. Square, rectangular, or round
columns are the most economical; use other shapes only when architectural requirements so dictate.
9.5.3 Walls
• Use the same wall thickness throughout a project if possible; this facilitates the reuse of equipment, ties,
and hardware. In addition, this minimizes the possibilities of error in the field. In all cases, maintain
sufficient wall thickness to permit proper placing and vibrating of concrete.
• Wall openings should be kept to a minimum number since they can be costly and time-consuming. A few
larger openings are more cost-effective than many smaller openings. Size and location should be
constant for maximum reuse of formwork.
• Brick ledges should be kept at a constant height with a minimum number of steps. Thickness as well as
height should be in dimensional units of lumber, approximately as closely as possible those of the
masonry to be placed. Brick ledge locations and dimensions should be detailed on the structural drawings.
• Footing elevations should be kept constant along any given wall if possible. This facilitates the use of
wall gang forms from footing to footing. If footing steps are required, use the minimum number possible.
• For buildings of moderate height, pilasters can be used to transfer column loads into the foundation walls.
Gang forms can be used more easily if the pilaster sides are splayed as shown in Fig. 9-9.
Figure 9-9 Pilasters
9.6 OVERALL STRUCTURAL ECONOMY
While it has been the primary purpose of this chapter to focus on those considerations that will significantly
impact the costs of the structural system relative to formwork requirements, the 10-step process below should
be followed during the preliminary and final design phases of the construction project as this will lead to
overall structural economy:
(1) Study the structure as a whole.
(2) Prepare freehand alternative sketches comparing all likely structural framing systems.
(3) Establish column locations as uniformly as possible, keeping orientation and size constant wherever possible.
9-9
x + 1”
x

(4) Determine preliminary member sizes from available design aids (see Section 1.8).
(5) Make cost comparisons based on sketches from Step 2 quickly, roughly, but with an adequate degree
of accuracy.
(6) Select the best balance between cost of structure and architectural/mechanical design considerations.
(7) Distribute prints of selected framing scheme to all design and building team members to reduce
unnecessary future changes.
(8) Plan your building. Visualize how forms would be constructed. Where possible, keep beams and columns
simple without haunches, brackets, widened ends or offsets. Standardize concrete sizes for maximum
reuse of forms.
(9) During final design, place most emphasis on those items having greatest financial impact on total
structural frame cost.
(10) Plan your specifications to minimize construction costs and time by including items such as early
stripping time for formwork and acceptable tolerances for finish.
Reference 9.4 should be consulted for additional information concerning formwork.
References
9.1 Concrete Buildings, New Formwork Perspectives, Ceco Industries, Inc., 1985.
9.2 Guide to Formwork for Concrete, ACI 347-04, American Concrete Institute, Farmington Hills, Michigan,
2004, 32 pp.
9.3 Concrete Floor Systems—Guide to Estimating and Economizing, SP041, Portland Cement Association,
Skokie, Illinois, 2000, 41 pp.
9.4 Hurd, M.K., Formwork for Concrete, (prepared under direction of ACI Committee 347, Formwork for
Concrete), SP-4, 6th Ed., American Concrete Institute, Detroit, Michigan, 1995.
9-10

10-1
Chapter 10
Fire Resistance
10.1 INTRODUCTION
State and municipal building codes throughout the country regulate the fire resistance of the various elements
and assemblies comprising a building structure. Structural frames (columns and beams), floor and roof
systems, and load bearing walls must be able to withstand the stresses and strains imposed by fully developed
fires and carry their own dead loads and superimposed loads without collapse.
Fire resistance ratings required of the various elements of construction by building codes are a measure of the
endurance needed to safeguard the structural stability of a building during the course of a fire and to prevent
the spread of fire to other parts of the building. The determination of fire rating requirements in building codes
is based on the expected fire severity (fuel loading) associated with the type of occupancy and the building
height and area.
In the design of structures, building code provisions for fire resistance are sometimes overlooked and this may
lead to costly mistakes. It is not uncommon, for instance, to find that a concrete slab in a waffle slab floor
system may only require a 3 to 4-1
/2
in. thickness to satisfy ACI 318 strength requirements. However, if the
building code specifies a 2-hour fire resistance rating for that particular floor system, the slab thickness may
need to be increased to 3-1
/2
to 5 in., depending on type of aggregate used in the concrete. Indeed, under such
circumstances and from the standpoint of economics, the fire-resistive requirements may indicate another
system of construction to be more appropriate, say, a pan-joist or flat slab/plate floor system. Simply stated,
structural members possessing the fire resistance prescribed in building codes may differ significantly in their
dimensional requirements from those predicated only on ACI 318 strength criteria. Building officials are
required to enforce the stricter provisions.
The purpose of this chapter is to make the reader aware of the importance of determining the fire resistance
requirements of the governing building code before proceeding with the structural design.
The field of fire technology is highly involved and complex and it is not the intent here to deal with the chem-
ical or physical characteristics of fire, nor with the behavior of structures in real fire situations. Rather, the goal
is to present some basic information as an aid to designers in establishing those fire protection features of con-
struction that may impact their structural design work.
The information given in this chapter is fundamental. Modern day designs, however, must deal with many
combinations of materials and it is not possible here to address all the intricacies of construction. Rational
methods of design for dealing with more involved fire resistance problems are available. For more
comprehensive discussions on the subject of the fire resistive qualities of concrete and for calculation methods
used in solving design problems related to fire integrity, the reader may consult Reference 10.1.

10.2 DEFINITIONS
Structural Concrete:
• Siliceous aggregate concrete: concrete made with normal weight aggregates consisting mainly of silica
or compounds other than calcium or magnesium carbonate.
• Carbonate aggregate concrete: concrete made with aggregates consisting mainly of calcium or magne-
sium carbonate, e.g., limestone or dolomite.
• Sand-lightweight concrete: concrete made with a combination of expanded clay, shale, slag, or slate or
sintered fly ash and natural sand. Its unit weight is generally between 105 and 120 pcf.
• Lightweight aggregate concrete: concrete made with aggregates of expanded clay, shale, slag, or slate
or sintered fly ash, and weighing 85 to 115 pcf.
Insulating Concrete:
• Cellular concrete: a lightweight insulating concrete made by mixing a preformed foam with Portland
cement slurry and having a dry unit weight of approximately 30 pcf.
• Perlite concrete: a lightweight insulating concrete having a dry unit weight of approximately 30 pcf
made with perlite concrete aggregate produced from volcanic rock that, when heated, expands to form a
glass-like material or cellular structure.
• Vermiculite concrete: a lightweight insulating concrete made with vermiculite concrete aggregate, a
laminated micaceous material produced by expanding the ore at high temperatures. When added to
Portland cement slurry the resulting concrete has a dry unit weight of approximately 30 pcf.
Miscellaneous Insulating Materials:
• Glass fiber board: fibrous glass roof insulation consisting of inorganic glass fibers formed into rigid
boards using a binder. The board has a top surface faced with asphalt and kraft reinforced with glass fibers.
• Mineral board: a rigid felted thermal insulation board consisting of either felted mineral fiber or cellular
beads of expanded aggregate formed into flat rectangular units.
10.3 FIRE RESISTANCE RATINGS
10.3.1 Fire Test Standards
The fire-resistive properties of building components and structural assemblies are determined by standard fire
test methods. The most widely used and nationally accepted test procedure is that developed by the American
Society of Testing and Materials (ASTM). It is designated as ASTM E 119 (Reference 10.2), Standard Methods
of Fire Tests of Building Construction and Materials. Other accepted standards, essentially alike, include the
National Fire Protection Association Standard Method No. 251; Underwriters Laboratories; U.L. 263;
10-2

American National Standards Institute’s ANSI A2-1; ULC-S101 from the Underwriters Laboratories of
Canada; and Uniform Building Code Standard No. 43-1.
10.3.2 ASTM E 119 Test Procedure
A standard fire test is conducted by placing an assembly in a test furnace. Floor and roof specimens are exposed
to controlled fire from beneath, beams from the bottom and sides, walls from one side, and columns from all
sides. The temperature is raised in the furnace over a given period of time in accordance with ASTM E 119
standard time-temperature curve shown in Fig. 10-1.
Figure 10-1 Standard Time-Temperature Relationship of Furnace Atmosphere (ASTM E 119)
This specified time-temperature relationship provides for a furnace temperature of 1000°F at five minutes from
the beginning of the test, 1300°F at 10 minutes, 1700°F at one hour, 1850°F at two hours, and 2000°F at four
hours. The end of the test is reached and the fire endurance of the specimen is established when any one of the
following conditions first occur:
(1) For walls, floors, and roof assemblies the temperature of the unexposed surface rises an average of 150°F
above its initial temperature of 325°F at any location. In addition, walls achieving a rating classification of
one hour or greater must withstand the impact, erosion and cooling effects of a hose steam test.
(2) Cotton waste placed on the unexposed side of a wall, floor, or roof system is ignited through cracks or
fissures developed in the specimen.
10-3
Chapter 10 • Design Considerations for Fire Resistance
2500
2000
1600
1000
500
0
0 2 4 6 8
Fire test time, hr.
Furnaceatmospheretemperature,°F

(3) The test assembly fails to sustain the applied load.
(4) For certain restrained and all unrestrained floors, roofs and beams, the reinforcing steel temperature rises
to 1100°F.
Though the complete requirements of ASTM E 119 and the conditions of acceptance are much too detailed for
inclusion in this chapter, experience shows that concrete floor/roof assemblies and walls usually fail by heat
transmission (item 1); and columns and beams by failure to sustain the applied loads (item 3), or by beam
reinforcement failing to meet the temperature criterion (item 4).
Fire rating requirements for structural assemblies may differ from code to code; therefore, it is advisable that
the designer take into account the building regulations having jurisdiction over the construction rather than
relying on general perceptions of accepted practice.
10.4 DESIGN CONSIDERATIONS FOR FIRE RESISTANCE
10.4.1 Properties of Concrete
Concrete is the most highly fire-resistive structural material used in construction. Nonetheless, the properties of
concrete and reinforcing steel change significantly at high temperatures. Strength and the modulus of elasticity are
reduced, the coefficient of expansion increases, and creep and stress relaxations are considerably higher.
Concrete strength, the main concern in uncontrolled fires, remains comparatively stable at temperatures
ranging up to 900°F for some concretes and 1200°F for others. Siliceous aggregate concrete, for instance, will
generally maintain its original compressive strength at temperatures up to 900°F, but can lose nearly 50% of
its original strength when the concrete reaches a temperature of about 1200°F. On the other hand, carbonate
aggregate and sand-lightweight concretes behave more favorably in fire, their compressive strengths remaining
relatively high at temperatures up to 1400°F, and diminishing rapidly thereafter. These data reflect fire test
results of specimens loaded in compression to 40% of their original compressive strength.
The temperatures stated above are the internal temperatures of the concrete and are not to be confused with the
heat intensity of the exposing fire. As an example, in testing a solid carbonate aggregate slab, the ASTM
standard fire exposure after 1 hour will be 1700°F, while the temperatures within the test specimen will vary
throughout the section: about 1225°F at 1
/4 in. from the exposed surface, 950°F at 3
/4 in., 800°F at 1 in., and
600°F at 1-1
/2 in.; all within the limit of strength stability.
It is to be realized that the strength loss in concrete subjected to intense fire is not uniform throughout the
structural member because of the time lag required for heat penetration and the resulting temperature gradients
occurring across the concrete section. The total residual strength in the member will usually provide an
acceptable margin of safety.
This characteristic is even more evident in massive concrete building components such as columns and girders.
Beams of normal weight concrete exposed to an ASTM E 119 fire test will, at two hours when the exposing
fire is at 1850°F, have internal temperatures of about 1200°F at 1 in. inside the beam faces and less than 1000°F
10-4

at 2 in. Obviously, the dimensionally larger concrete sections found in main framing systems will suffer far less
net loss of strength (measured as a percentage of total cross-sectional area) than will lighter assemblies.
Because of the variable complexities and the unknowns of dealing with the structural behavior of buildings
under fire as total multidimensional systems, building codes continue to specify minimum acceptable levels of
fire endurance on a component by component basis—roof/floor assemblies, walls, columns, etc. It is known,
for instance, that in a multi-bay building, an interior bay of a cast-in-place concrete floor system subjected to
fire will be restrained in its thermal expansion by the unheated surrounding construction. Such restraint
increases the structural fire endurance of the exposed assembly by placing the heated concrete in compression.
The restraining forces developed are large and, under elastic behavior, would cause the concrete to exceed its
original compressive strength were it not for stress relaxations that occur at high temperatures. According to
information provided in Appendix X3 of ASTM E 119, cast-in-place beams and slab systems are generally
considered restrained (see Table 10-5 in Section 10.4.3).
In addition to the minimum acceptable limits given in the building codes, the use of calculation methods for
determining fire endurance are also accepted, depending on the local code adoptions (see Reference 10.1 and 10.3).
10.4.2 Thickness Requirements
Test findings show that fire resistance in concrete structures will vary in relation to the type of aggregate used.
The differences are shown in Table 10-1 and 10-2.
Table 10-1 Minimum Thickness for Floor and Roof Slabs and Cast-in-Place Walls, in.
(Load-Bearing and Nonload-Bearing)—(Reference 10.4)
Table 10-2 Minimum Concrete Column Dimensions, in.—(Reference 10.4)
In studying the tables above it is readily apparent that there may be economic benefits to be gained from the
selection of the type of concrete to be used in construction. The designer is encouraged to evaluate the alternatives.
10.4.3 Cover Requirements
Another factor to be considered in complying with fire-resistive requirements is the minimum thickness of
concrete cover for the reinforcement. The concrete protection specified in ACI 318 for cast-in-place concrete
10-5

will generally equal or exceed the minimum cover requirements shown in the following tables, but there are a
few exceptions at the higher fire ratings and these should be noted.
The minimum thickness of concrete cover to the positive moment reinforcement is given in Table 10-3 for one-
way or two-way slabs with flat undersurfaces.
The minimum thickness of concrete cover to the positive moment reinforcement (bottom steel) in reinforced
concrete beams is shown in Table 10-4.
Table 10-3 Minimum Cover for Reinforced Concrete Floor or Roof Slabs, in.—(Reference 10.4)
Table 10-4 Minimum Cover to Main Reinforcing Bars in Reinforced Concrete Beams, in.
(Applicable to All Types of Structural Concrete)
The minimum cover to main longitudinal reinforcement in columns is shown in Table 10-6.
10-6
Fire-Resistance Rating (hours)
Restrained Unrestrained
Concrete Aggregate Type 1 11/2 2 3 4 1 11/2 2 3 4
Siliceous 3/4 3/4 3/4 3/4 3/4 3/4 3/4 1 11/4 15/8
Carbonate 3/4 3/4 3/4 3/4 3/4 3/4 3/4 3/4 11/4 11/4
Sand-lightweight or lightweight 3/4 3/4 3/4 3/4 3/4 3/4 3/4 3/4 11/4 11/4

Table 10-5 Construction Classification, Restrained and Unrestrained
(Table X3.1 from ASTM E 119)
Table 10-6 Minimum Cover for Reinforced Concrete Columns, in.
10-7
I. Wall bearing
Single span and simply supported end spans of multiple bays:A
(1) Open-web steel joists or steel beams, supporting concrete slab, precast units, or metal decking unrestrained
(2) Concrete slabs, precast units, or metal decking unrestrained
Interior spans of multiple bays:
(1) Open-web steel joists, steel beams or metal decking, supporting continuous concrete slab restrained
(2) Open-web steel joists or steel beams, supporting precast units or metal decking unrestrained
(3) Cast-in-place concrete slab systems restrained
(4) Precast concrete where the potential thermal expansion is resisted by adjacent constructionB
restrained
II. Steel framing:
(1) Steel beams welded, riveted, or bolted to the framing members restrained
(2) All types of cast-in-place floor and roof systems (such as beam-and-slabs, flat slabs, pan joists,
and waffle slabs)where the floor or roof system is secured to the framing members
restrained
(3) All types of prefabricated floor or roof systems where the structural members are secured to
the framing members and the potential thermal expansion of the floor or roof system is resisted
by the framing system or the adjoining floor or roof constructionB
restrained
III. Concrete framing:
(1) Beams securely fastened to the framing members restrained
(2) All types of cast-in-place floor or roof systems (such as beam-and-slabs, pan joists, and waffle
slabs) where the floor system is cast with the framing members
restrained
(3) Interior and exterior spans of precast systems with cast-in-place joints resulting in restraint
equivalent to that which would exist in condition III (1)
restrained
(4) All types of prefabricated floor or roof systems where the structural members are secured to
such systems and the potential thermal expansion of the floor or roof systems is resisted by the
framing system or the adjoining floor or roof constructionB
restrained
IV. Wood construction:
All Types unrestrained
A
Floor and roof systems can be considered restrained when they are tied into walls with or without tie beams, the walls
being designed and detailed to resist thermal thrust from the floor or roof system.
B
For example, resistance to potential thermal expansion is considered to be achieved when:
(1) Continuous structural concrete topping is used,
(2) The space between the ends of precast units or between the ends of units and the vertical face of supports is filled
with concrete or mortar, or
(3) The space between the ends of precast units and the vertical faces of supports, or between the ends of solid or
hollow core slab units does not exceed 0.25% of the length for normal weight concrete members or 0.1% of the
length for structural lightweight concrete members.
*Copyright ASTM. Reprinted with permission.

10.5 MULTICOURSE FLOORS AND ROOFS
Symbols: Carb = carbonate aggregate concrete
Sil = siliceous aggregate concrete
SLW = sand-lightweight concrete
10.5.1 Two-Course Concrete Floors
Figure 10-2 gives information on the fire resistance ratings of floors that consist of a base slab of concrete with
a topping (overlay) of a different type of concrete.
Figure 10-2 Fire Resistance Ratings for Two-Course Floor Slabs
10-8
5
4
3
2
1
5
4
3
2
1
5
4
3
2
1
5
4
3
2
1
0 1 2 3 4 50 1 2 3 4 5
0 1 2 3 4 5 0 1 2 3 4 5
Sil baseCarb base
Thickness of normal weight concrete base slab, in.
Thickness of sand-lightweight concrete base slab, in.
Carboverlay
Overlaythickness,in.
Siloverlay
Thicknessofsand-lightweight
concreteoverlay,in.
4 hr.
4 hr. 4 hr.
4 hr.
3
1
2
3
1
2
3
1
2
3
1
2

10.5.2 Two-Course Concrete Roofs
Figure 10-3 gives information on the fire resistance ratings of roofs that consist of a base slab of concrete with
a topping (overlay) of an insulating concrete; the topping does not include built-up roofing. For the transfer of
heat, three-ply built-up roofing contributes 10 minutes to the fire resistance ratings; thus, 10 minutes may be
added to the values shown in the figure.
10.5.3 Concrete Roofs with Other Insulating Materials
Figure 10-4 gives information on the fire resistance ratings of roofs that consist of a base slab of concrete with
an insulating board overlay; the overlay includes standard 3-ply built-up roofing.
Figure 10-3 Fire Resistance Ratings for Two-Course Roof Slabs
10-9
Cellular concrete
Concrete
Vermiculite concrete
Concrete
Perlite concrete
Concrete
Thickness of concrete base slab, in.
Thicknessofcellular
concreteoverlay,in.
Thicknessofperlite
concreteoverlay,in.
Thicknessofvermiculite
concreteoverlay,in.
Carb base Sil base SLW base
3
2
1
3
2
1
3
2
1
0 2 4 0 2 4 0 2 4
4
3
2
1
4
3
2
1
4
3
2
1
4
3
2
1
4
3
2
1
4
3
2
1
0 2 4 0 2 4 0 2 4
0 1 2 3 4 0 1 2 3 4 0 1 2 3 4
4 hr.
4 hr.4 hr.
4 hr.
4 hr.4 hr.
3
2
1
3
2
1
3
2
1
3
2
1
3
2
1
3
2
1
3
2
1
4 hr.
3
2
1
4 hr.
3
2
1
4 hr.
3
2
1

Figure 10-4 Fire Resistance Ratings for Roof Slabs with Insulating Overlays and
Standard 3-Ply Built-Up Roofing
Reference
10.1 Reinforced Concrete Fire Resistance, Concrete Reinforcing Steel Institute, Schaumburg, Illinois, 256 pp.
10.2 Standard Test Methods for Fire Test of Buildings Construction and Materials-E119-00a, American Society
of Testing and Materials, 2000.
10.3 Code Requirements for Determining Fire Resistance of Concrete and Masonry Construction Assemblies,
ACI/TMS Standard, ACI 216.1-07/TMS-0216-07.
10.4 International Building Code, International Code Council, 2009.
10-10
Concrete
Concrete
Mineral board
Glass fiber board
Standard 3-ply
built-up roofing
Standard 3-ply
built-up roofing
Thicknessof
mineralboard,in.
Thicknessof
glassfiberboard,in.
3
2
1
0
3
2
1
0
3
2
1
0
3
2
1
0
3
2
1
0
3
2
1
0
1 2 3 4 1 2 3 4 1 2 3 4
1 3 5 1 3 5 1 3 5
4 hr.
3
2
1
4 hr.
3
2
1
4 hr.
3
2
1
4 hr.
3
2
1
4 hr.
3
2
1
4 hr.
3
2
1

11-1
Chapter 11
Earthquake Forces
11.1 INTRODUCTION
The objective of this chapter is to introduce the basic seismic design provisions of the International Building
Code (2009 IBC) that apply to the structures intended to be within the scope of this publication where seismic
forces are resisted entirely by moment frame or shearwalls. The 2009 IBC refers or adopts with modification
provisions from ASCE 7-052
. Reference to both documents will be made throughout this chapter as applicable.
The material in this chapter does not cover structures with horizontal or vertical irregularities (ASCE 12.3.2)
and assumes rigid diaphragm typical for cast-in-place concrete floor systems. Examples of structural systems
with horizontal or vertical irregularities are shown in Figures 11-1 and 11-2 respectively. For comprehensive
background on seismic design and detailing requirements for all cases, refer to References 11.1, 11.2 and 11.3.
The ASCE-7 contains contour maps for the maximum considered earthquake (MCE) spectral response
accelerations (5 percent of critical damping) at periods of 0.2 second (Ss
) and 1.0 second (S1
). The mapped
values of Ss
and S1
are based on Site Class B (see below for site class definition). The earthquake effects that
buildings and structures are proportioned to resist are based on what is called Design Basis Earthquake (DBE).
The design seismic forces prescribed in the ASCE-7 are generally less than the elastic inertia forces induced
by the DBE (Reference 11.3). Figure 11-3 shows the relation between the force generated by the DBE if the
structure were to be designed for elastic response for this force (Ve
) and the actual Code prescribed force used
in design (V). The Figure also shows the inelastic response under the Code prescribed force (V). Structures
subjected to seismic forces must resist collapse when subjected to several cycles of loading in the inelastic
range. Therefore, critical regions of certain members must be designed and detailed to safely undergo sufficient
inelastic deformability. The building code contains structural detailing requirements to enable the structure and
members to dissipate seismic energy by inelastic deformation in order to prevent collapse.
11.2 SEISMIC DESIGN CATEGORY (SDC)
The ASCE 7-05 requires that a Seismic Design Category (SDC) (A, B, C, D, E, or F) be assigned to each
structure (11.6). A SDC is used to determine:
(1) Permissible structural systems
(2) Level of detailing
(3) Limitations on height and irregularity
(4) The components of the structure that must be designed and detailed for seismic resistance

11-2
Torsional irregularity
Diaphragm
discontinuityOpening
Re-entrant
corner
Plan irregularities
Out-of-plane Vertical Element Offsets Nonparallel Systems
Figure 11-1 Structures with Horizontal Irregularities
Figure 11-2 Structures with Vertical Irregularities
Vertical Stiffness Irregularity—Soft Story Weight (Mass) Irregularity
Vertical Geometric Irregularity Vertical Strength Irregularity—Weak Story
Soft story Soft story
Shearwall or braced frame
Heavy
mass
Weak story

11-3
Chapter 11 • Design Considerations for Earthquake Forces
Force
Forcefromdesignearthquake
(Elasticresponse)
Ve
BaseshearV=Ve/(R/I)
Deformation
V=Designlateralforce
δxe=Lateraldeflectionfromelastic
analysis
δxe=Lateraldeflectionconsideringthe
inelasticresponse
Cd=Deflectionamplificationfactor
ASCE12.2-1
I=ImportancefactorASCE11.5.1
R=Responsemodificationcoefficient
ASCE12.2-1
δxe
δx=Cdδxe/I
Figure11-3CodeDesignSeismicForcesComparedtoDBEEffect

11-4
(5) Structures exempt from seismic design requirements
(6) The types of lateral force analysis that must be performed
The SDC is a function of two parameters 1) the Occupancy Category I, II, or III (ASCE 11.5), which depends
on the occupancy and use of the building, and 2) the design spectral response acceleration. The following
procedure summarizes the determination of the SDC:
(1) From 2009 IBC Figures 1613.5(1) through 1613.5(14), determine the maximum considered earthquake
spectral response acceleration at short period (0.2 second) Ss
and at one second and S1
. These values can
alternatively be obtained from the website of the United States Geological Survey (USGS) by inputting the
zip code of the building location.
(2) Depending on the soil properties at the site, determine the site class definition (A, B, C, D, E, or F), from
1613.5.2. In the absence of sufficient details on site soil properties it is allowed to assume site class D
unless the building official specifies higher site class (E or F).
(3) Based on the magnitude of the maximum considered earthquake spectral response accelerations (Ss
and S1
)
from step 1 and the site class from step 2 determine the site coefficients Fa
and Fv
from Tables 1613.5.3(1)
and 1613.5.3(2). Figures 11-4 and 11-5 can be used to determine Fa
and Fv
(4) Calculate the adjusted (for the site class effects) maximum considered earthquake spectral response
acceleration SMS
and SM1
as follows:
SMS
= Fa
Ss
SM1
= Fv
S1
(5) Calculate the five percent damped design spectral response acceleration at short period SDS
, and at one
second period SD1
(1613.5.4):
SDS
= 2/3 SMS
SD1
= 2/3 SM1
(6) SDC (A, B, C, D, E, or F) is determined based on the Occupancy Category (I, II, or III) as defined in 1602
and the short period design spectral response accelerations SDS. Another seismic design category is
determined based on the occupancy category and SD1 (2009 IBC Table 1613.5.6(1) and Table 1613.5.6(2)).
The more critical of the two categories shall be used. Figure 11-6 sumaries how to determine the seismic
design category based on the design response accelerations SD1 and SDS
The 2009 IBC allows the SDC to be determined based on SDS alone (from table 1613.5.6(1)) provided that all
the following conditions apply (1613.6.1):
a) The approximate fundamental period of the structure, T (see Section 11.6.1 for T calculations), in each
of the two orthogonal directions is less than 0.8 Ts
where Ts
= SD1
/SDS.
b) The seismic response coefficient, Cs
(see Section 11.6.1) is determined based on SDS.

11-5
Soil E
Soil D
Soil C
Soil B
Soil A
0.25 0.5 0.75 1 1.25
SiteCoefficientFa
2.5
2
1.5
1
0.5
Response Accelaration for Short Period Ss
Figure 11-4 Site Coefficient Fa
0.1 0.2 0.3 0.4 0.5
SiteCoefficientFv
3.5
3
2.5
2
1.5
1
0.5
Response Accelaration at 1 sec Period S1
Figure 11-5 Site Coefficient Fv

11-6
c) The diaphragms are rigid as defined in 2009 IBC Section 1602.
Applicable to buildings of moderate height, this exception can make a substantial difference in the SDC of the
building which impacts the type of detailing required and ultimately plays a key role in the building economy.
The utilization of this exception is illustrated in the examples given in this chapter.
I
II
III
IV
A
B C
C
D
Occupancycategory
0.067
0.133
0.2
0.167
0.33
0.5
SD1
SDS
Structures located where S1 ≥ 0.75 are assigned SDC E if Occupancy Catagory I, II, III, and SDC F if
Occupancy Category IV.
The SDC is the more severe in accordance to SD1 or SDS
Where S1 0.75, the Seismic Design Category is permitted to be determined base on SDS alone,
if T 0.8Ts where Ts = SD1/SDS (see ASCE 11.6 for other limitations)
Figure 11-6 Seismic Design Category Based on SDS and SD1
11.3 REINFORCED CONCRETE EARTHQUAKE-RESISTING STRUCTURAL SYSTEMS
The basic reinforced concrete seismic force resisting systems are shown in Figure 11-7. The permitted
structural system, height limitations, and reinforcement detailing depend on the determined SDC. A brief
description of each system follows:
(1) Bearing Wall System: Load bearing walls provide support for most or all of the gravity loads. Resistance
to lateral forces is provided by the same walls acting as shearwalls.
(2) Building Frame System: A structural system, with essentially a complete space, frame provides support
for the gravity loads. Resistance to lateral forces is provided by shearwalls.
(3) Moment-Resisting Frame System: An essentially complete space frame provides support for the gravity
loads and resistance to lateral loads at the same time (simultaneously).

11-7
Bearing Wall System Building Frame System
Moment-Resisting Frame System Dual System
Figure 11-7 Earthquake-Resisting Structural Systems of Reinforced Concrete
(4) Dual System: A structural system with space frames and walls to provide support for the gravity loads.
Resistance to earthquake loads is provided by shearwalls and moment-resisting frames. The shearwalls and
moment-resisting frames are designed to resist the design base shear in proportion to their relative rigidities.
The moment resisting frame must be capable of resisting at least 25 percent of the design base shear (1617.6.1).
Based on ASCE 7, Table 12.2-1 shows the permitted system and height limitation for each SDC. The table also
shows the values of R and Cd. The table could be used to select the required seismic-force-resisting systems
for a specific SDC. The building frame system and the moment-resisting frame system are commonly used and
are suitable for buildings of regular shape and moderate height. Bearing wall and dual system buildings are,
therefore, not discussed in this publication.
11.4 STRUCTURES EXEMPT FROM SEISMIC DESIGN REQUIREMENTS
The 2009 IBC allows certain structures to be exempt from the seismic design requirements (1613.1). These
structures include: detached one- and two-family dwellings with SDC A, B, or C or located where the mapped
short-period spectral response acceleration Ss is less than 0.4g. Agricultural storage structures intended only
for incidental human occupancy are also included in the exemption.
Also structures located at the following locations are assigned SDC A:
a) Where SS ≤ 0.15g and S1 ≤ 0.04g
b) Where SDS ≤ 0.167g and SD1 ≤ 0.067g

11-8
g y ( )
A B C D E F
160 ft 160 ft 100 ft 5 5
4 4
160 ft 160 ft 100 ft 6 5
5 4.5
8 5.5
5 4.5
3 2.5
Special
Shearwall
7 5.5
Ordinary
Shearwall
6 5
Special
Shearwall
160 ft 100 ft 100 ft 6.5 5
Ordinary
Shearwall
5.5 4.5
Permitted with the indicated height limit (if any)
Not permitted
CdR
SDC
With
Intermediate
Moment Frame
Intermediate Frame
Ordinary Frame
Special Reinforced Concrete
Shearwall
Ordinary Reinforced Concrete
Shearwall
Basic Seismic Force Resisting System
Moment Resisting Frame
Dual System
Special Reinforced Concrete
Shearwall
Ordinary Reinforced Concrete
Shearwall
Special Frame
With Special
Moment Frame
Bearing Wall
Building Frame
Table 11-1 Permitted Building Systems for Different SDC (ASCE Table 12.2-1)
11.5 EARTHQUAKE FORCES
The IBC code requires a structural analysis procedure to determine the magnitude and distribution of
earthquake forces. The permitted analysis procedure for certain structures depends on the seismic design
category and the structure characteristics (Occupancy Catagory, structural irregularity, height, and location).
The 2009 IBC refers to ASCE 7-0511.2 (Minimum Design Loads for Buildings and other Structures) for the
permitted analysis procedures for different structures (ASCE Table 12.6.1). Therefore, the provisions of
ASCE 7-05 will be used from here on. For structures assigned to SDC A, the structure must be analyzed for
the effects of static lateral forces applied independently in each of two orthogonal directions. In each direction,
the static lateral forces at all levels must be applied simultaneously. The lateral force considered at each floor
level is equal to one percent (1%) of the portion of the total dead load of the structure, D, assigned to that level.

11-9
For structures assigned to seismic design categories B, C, D, E or F, the ASCE introduces three analytical
procedures to account for earthquake effects; equivalent lateral force analysis (ASCE 12.8), modal response
spectrum analysis (ASCE 12.9) and seismic response history procedure (ASCE Chapter 16). For buildings
within the scope of this publication, the equivalent lateral forces procedure provides the most suitable
approach. The following sections present the equivalent lateral force method. For other analytical procedures
References 11.1 and 11.2 should be consulted.
11.6 EQUIVALENT LATERAL FORCE PROCEDURE
For reinforced concrete structures with or without irregularities and assigned to SDC B or C, the equivalent lat-
eral force procedure can be used. Also the equivalent lateral force method can be used for regular structures
assigned to SDC D, E, or F provided that the fundamental period of the structure T 3.5 Ts
where Ts
= SD1
/SDS
(see below for calculation of T). Limitations on the applicability of this method to irregular structures assigned
to SDC D, E, or F are given in ASCE (Table 12.6-1).
11.6.1 Design Base Shear
The seismic base shear, V, in a given direction is a fraction of the dead weight of the structure. V is calculated
from:
V = Cs
W
where:
Cs
= seismic response coefficient
W = the effective seismic weight of the structure which includes the total dead load and the loads listed
below (ASCE 12.7.2):
(1) In areas used for storage, a minimum of 25% of the floor live load (floor live load in public garages and
open parking structures need not be included)
(2) Where an allowance for partition load is included in the floor load design, the actual partition weight or a
minimum weight of 10 psf of floor area, whichever is greater
(3) Total operating weight of permanent equipment
(4) 20% of flat roof snow load where flat roof snow load exceeds 30 psf
The seismic response coefficient is calculated as follows:
For T ≤ TL
For T TLCs
=
SD1
TL
T2 R
I
⎛
⎝
⎜
⎞
⎠
⎟
Cs
=
SD1
T
R
I
⎛
⎝
⎜
⎞
⎠
⎟
≤
SDS
R
I
⎛
⎝
⎜
⎞
⎠
⎟

11-10
The value of Cs
should not be taken less than 0.01. For locations where S1
≥ 0.6g Cs
should not be less than:
where:
TL
= long period transition period (ASCE Figure 22-15 to Figure 22-20). It is important to point out that TL
is the transition between the constant velocity and constant displacement in the response spectra for certain
location. This should have no effect on buildings with low to moderate heights. In effect the above first
equation for Cs
calculations controls the analysis of buildings addressed in this publication.
R = response modification factor depending on the basic seismic-force-resisting system, from ASCE
Table 12.2-1. Table 11-1 lists the values for R for different reinforced concrete seismic force resisting
systems.
I = importance factor depending on the nature of occupancy, from ASCE 11.5.1
T = the fundamental period of the structure in seconds. T can be calculated from the following equation:
T = Ct
(hn
)x
where:
Ct
= building period coefficient
Ct
= 0.016 for concrete moment resisting frames
Ct
= 0.02 for other concrete systems
hn
= building height in feet
x = 0.9 for concrete moment resisting frames
x = 0.75 for other concrete systems
Figure 11-8 shows the fundamental period for concrete building systems for different heights.
For concrete moment resisting frame buildings with less than 12 floors and story height of 10 feet minimum,
T calculations can be further simplified to T = 0.1 N where N is the number of stories
For concrete shearwall structures the fundamental period can be approximated using ASCE Eq. 9.5.5.3 2-2
11.6.2 Vertical Distribution of Seismic Forces
In the equivalent lateral force method the design base shear V is distributed at different floor levels as follows
(ASCE 12.8.3):
Cs
≥
0.5S1
R
I
⎛
⎝
⎜
⎞
⎠
⎟

11-11
Moment resisting frame system
Other concrete systems
FundamentalPeriodT,Sec.
1.1
1
0.9
0.8
0.7
0.6
0.5
0.4
0.3
0.2
0.1
0
Building Height h, ft.
10 20 30 40 50 60 70 80 90 100
Fx
= Cvx
V
where:
Fx = lateral force at floor level x
Cvx= vertical distribution factor for level x
k = a distribution exponent related to the building period T
= 1 for T ≤ 0.5 second
= 2 for T ≥ 2.5 second
= 2 for 0.5 second T 2.5 second, or linear interpolation between 1 and 2
Cvx
=
wx hx
k
wihi
k
i=1
n
∑
Figure 11-8 Fundamental Period for Reinforced Concrete Buildings

11-12
hi
and hx
= the height in feet from the base level to i or x
wi
and wx
= the portion of W assigned to level i or x
For most of the structures covered in this publication, the fundamental period T is less than 0.5 second.
For this case the above equation simplifies to:
11.6.2.1 Distribution of Seismic Forces to Vertical Elements of the Lateral Force
Resisting System
The seismic design story shear Vx
in any story x is the sum of the lateral forces acting at that story in addition
to the lateral forces acting on all the floor levels above (ASCE Eq. 12.8-13):
Figure 11-9 shows the vertical distribution of the seismic force Fx
and the story shear Vx
in buildings with T ≤ 0.5.
The lateral shear force Vx
is typically transferred to the lateral force resisting elements (shearwalls or frames)
by the roof and floors acting as diaphragms. At each level the floor diaphragm distributes the lateral forces from
above to the shearwalls and frames below. The distribution of the lateral force to the lateral force resisting
elements (shearwalls or frames) depends on the relative rigidity of the diaphragm and the lateral force resisting
elements. For analysis purposes the diaphragms are typically classified as rigid, semi-rigid, and flexible.
Cast-in-place concrete floor systems are considered and modeled as rigid diaphragms. In rigid diaphragms, the
lateral force Vx
is distributed to the shearwalls and frames in proportion to their relative stiffnesses.
Figure 11-9 Vertical Distribution of Seismic Base Shear in Low-rise Buildings (T ≤ 0.5 sec)
For building frame system (consisting of shearwalls and frames) the shearwalls are designed to resist the entire
story shear Vx. For SDC D, E, and F the frames must be designed to resist the effects caused by the lateral
deflections, since they are connected to the walls through the floor slab (ASCE 12.2.4).
The seismic design story shear Vx is considered to act at the center of mass of the story. The center of mass is
the location where the mass of an entire story may be assumed to be concentrated. The location of the center
of mass can be determined by taking the moment of the components weights about two orthogonal axes x and y.
The distribution of Vx to the walls and frames depends on the relative location of the center of mass with
Vx = Fi
i= x
n
∑
Cvx
=
wx
hx
wi
hi
i=1
n
∑
Force Shear
Fx
Vx

11-13
CM,CR
V
(a) Centers of mass and rigidity at the same point
(b) Centers of mass and rigidity at distinct points
Mt = Ve
CR CM
V V
e
respect to the location of the center of rigidity of the floor. The center of rigidity is the point where the
equivalent lateral story stiffness (for frames or walls) may be considered to be located.
When the centers of mass and rigidity coincide, the lateral load resisting elements in the story displace an equal
distance horizontally (translate) and the story shear Vx
is distributed to the lateral force resisting elements in
proportion to their relative stiffnesses. If the center of mass does not coincide with the center of rigidity the
lateral load resisting elements in the story displace unequally (translate and rotate) and the story shear Vx
is
distributed to the lateral force resisting elements depending on their location and their relative stiffnesses.
Figure 11-10 illustrates the two cases. ASCE 12.8.4.2 requires that the horizontal force acting on each element
be increased due to accidental torsional moment. Such moment is the result of an assumed offset between the
center of mass and the center of rigidity of the seismic force resisting elements. The assumed offset is
5 percent of the building plan dimension perpendicular to the force direction at each level and is intended to
account for inaccuracies in building weight and stiffness estimate.
Figure 11-10 Floor Displacements

11-14
The location of the center of rigidity for a story can be determined by calculating the coordinates and from
arbitrary located origin as follows:
where:
(ki
)y
= lateral stiffness of lateral load resisting element (wall or frame) in the y-direction
(ki
)x
= lateral stiffness of lateral load resisting element (wall or frame) in the x-direction
xi
and yi
= coordinates measured from arbitrary located origin to the centroid of lateral force
resisting element i.
The distribution of the seismic shear force Vx
(at floor x) to the different lateral force resisting element
(shearwalls or frames) can be calculated from the following equations:
For force seismic force Vx
applied in x direction
For force seismic force Vx
applied in y direction
where:
xi
, yi
= perpendicular distances from the lateral force resisting element i to the center of
rigidity parallel to x and y axes respectively
Jr = rotational stiffness for all lateral force resisting elements in the story
ex , ey = perpendicular distance from the center of mass to the center of rigidity or assumed
eccentricities parallel to x and y axes respectively
(ki)x , (ki)y = stiffnesses of lateral force resisting element in x and y direction, respectively
11.6.2.2 Direction of Seismic Load
To determine the seismic force effects on different structural members, a structural analysis for the building needs
to be preformed. The design seismic forces should be applied in the direction which produces the most critical load
effect in each structural component. Provisions on application of loading are given in ASCE 12.5 as a function of
the SDC and the irregularity of the structure. Applications of these provisions for irregular structures are beyond
the scope of this publication.
xr
yr
xr
=
ki( )∑ y
xi
ki( )y∑
yr =
ki( )∑ x
yi
ki( )x∑
= xi
2
∑ ki( )y
+ yi
2
∑ ki( )x
Vi( )y
=
ki( )y
ki( )x∑
Vx +
yi
ki( )y
Jr
Vx ex
Vi( )x
=
ki( )x
ki( )x∑
Vx
+
yi
ki( )x
Jr
Vx
ey

11-15
11.6.3 Load Combinations for Seismic Design
The seismic forces effect E in the load combinations introduced in Chapter 2 (Section 2.2.3) is the combined
effect of horizontal and vertical earthquake induced forces and is calculated as follows:
For load combination ACI Equation 9-5:
E = ρQE + 0.2 SDSD
For load combination ACI Equation 9-7:
E = ρQE - 0.2 SDSD
where ρ is a redundancy factor based on the structural redundancy present in the building. For SDC A, B, or
C, ρ = 1. For structures assigned SDC D, E or F, ρ = 1.3, unless the conditions in ASCE 12.2.4.2 are met, in
such case it is permitted to assumed ρ = 1. Also, for drift calculation and P-delta effects ρ = 1 for all seismic
design categories (ASCE 12.3.4.1).
11.7 OVERTURNING
A building must be designed to resist the overturning effects caused by the seismic forces (ASCE 12.8.5).
The overturning moment (Mx) at any level x is determined from the following equation:
where
Fi = the portion of the base shear V, induced at level i.
hi
and hx
= the height in feet from the base to level i or x.
11.8 STORY DRIFT
ASCE 7-05 specifies maximum allowable limits for story drift ⌬ resulting from the design earthquake (ASCE
Table 12.12-1). Drift control is important to limit damage to partitions, shafts and stair enclosures, glass and
other fragile nonstructural elements. The design story drift ⌬ is the difference of the lateral deflection δx
(resulting from the design earthquake) at the floor level x at the top and bottom of story under consideration
(Figure 11-11). The design story drift ⌬x
is calculated as follows:
⌬x
= δx
- δx−1
where δx
and δx-1
are the magnified lateral displacement at the top and bottom of the story considered
(Figure 11-11). The magnified lateral displacement is calculated from the following equation:
δx
= Cd
δxe
/I
Mx = Fi hi − hx( )
i= x
n
∑

11-16
where
Cd
= deflection amplification factor presented in ASCE Table 9.5.2.2
δxe
= elastic lateral deflection (in.) due to the code prescribed seismic forces
I = occupancy importance factor defined in 11.6.1
Figure 11-11 Interstory Drift, ⌬
11.9 P-⌬ EFFECT
Seismic forces cause the structure to deflect laterally. As a result, secondary moments are induced in the
structural members due to the displaced gravity load as shown in Figure 11-12. This secondary moment
effect is known as the P-⌬ effect. P-⌬ effects are not required to be considered if the stability index θ is equal
to or less than 0.10. The stability index θ is calculated as follows (ASCE Eq. 12.8-16):
where:
Px = total unfactored vertical design load at and above level x (kips)
⌬ = design story drift (inches) occurring simultaneously with Vx.
Vx
= seismic shear force (kips) acting between level x and x-1
hsx
= story height (feet) below level x
Cd
= deflection amplification factor defined in 11.8
The stability coefficient, θ, shall not exceed θmax
calculated as follows:
where:
β = the ratio of shear demand to shear capacity for the story between level x and x-1 which may
be conservatively taken equal to 1.0.
θmax
=
0.5
βCd
≤ 0.25
θ =
Px
Δ
Vx
hsx
Cd
δx
δx-1

11-17
For cases when the stability index θ is greater than 0.10 but less than or equal to θmax
, the drift and element forces shall
be calculated including P-⌬ effects. To obtain the story drift including the P-⌬ effect, the design story drift shall
be multiplied by 1.0/(1.0 - θ). Where θ is greater than θmax
the structure is potentially unstable and shall be
redesigned to provide the needed stiffness to control drift.
Vx
Px
Level x+1
Level x
hsx
Level x-1
Figure 11-12 P-⌬ Effects
11.10 DESIGN AND DETAILING REQUIREMENTS
The magnitudes of the design seismic forces determined by the analysis procedure (Equivalent Lateral Force
Procedure) are reduced from the magnitudes of the actual forces that an elastic structure may experience during
an earthquake by the response modification factor R (see Figure 11-3). It is uneconomical and unnecessary to
design a structure to respond elastically when subjected to the anticipated ground motion resulting from an
earthquake. Traditionally, structures and their components are designed to yield under the code prescribed
seismic forces. However, the yielding members are expected to undergo substantial additional deformation
beyond the yield point while retaining strength capacity. This demonstrates a minimum level of ductility to
prevent collapse that is suitable for the seismic design category assigned to the structure. In addition to
proportioning the structural members’ dimensions and reinforcement for the seismic force effects, structures
must be properly detailed so that they are able to dissipate the earthquake energy through inelastic
deformation and provide the required ductility. The IBC requires compliance with the requirements of the
ACI 318 Code for design of reinforced concrete structures to achieve the required ductility. Design and
detailing requirements for ordinary structure members are presented in Chapters 1 through 19 in the ACI 318
Building Code and are as given in this book. No additional requirements are required for ordinary shear walls
and ordinary moment resisting frames assigned to SDC A. Additional requirements for the design and detailing
for special reinforced concrete shear walls, intermediate moment frame, and special moment frame are
presented in Chapter 21 in the ACI 318 Building Code. Table 11-2 presents a summary of ACI 318-08 sections
need to be satisfied for different cast-in-place concrete frames and walls.

11-18
Table 11-2 ACI Detailing Requirements for Seismic Design
11.11 EXAMPLES
Two examples are provided to illustrate the application of the equivalent lateral force procedure. The two
examples provide design calculations for shearwalls (Example 1) and typical frame flexural member (Example
2). For comprehensive coverage for other seismic design requirements see Reference 11.3.
11.11.1 Example 1 – Building # 2 Alternate (2) Shearwalls
Seismic design data
Assuming that the building is located in the Midwest with the maximum considered earthquake spectral
response accelerations for short period (0.2 second) and one second period determined from IBC maps
(2009 IBC Figures 1613.5(1) through 1613.5(14)) as follows:
SS
= 0.26 g
S1
= 0.12 g
Soil site class definition is D
Occupancy category I
Structural System Section
(a) Ordinary moment frames 21.2
(b) Ordinary RC structural walls N/A
(c) Intermediate moment frames 21.3
(d) Intermediate precast walls 21.4
(e) Special moment frames 21.1.3—21.1.7 21.5 21.8
(f) Special structural walls 21.1.3—21.1.7 21.9
(g) Special structural walls constructed using precast concrete 21.1.3—21.1.7 21.10

11-19
Determination of the Seismic Design Category SDC
(1) Based on the values of Ss
and S1
and site class D, determine the site coefficients Fa
and Fv
from
ASCE Tables 11.4-1 and 114.2. Notice that linear interpolation is performed.
Fa
= 1.59
Fv
= 2.32
(2) Calculate the adjusted maximum considered earthquake spectral response acceleration:
SMS
= Fa
Ss
= 1.59 ϫ 0.26 = 0.41g
SM1
= Fv
S1
= 2.32 ϫ 0.12 = 0.28g
(3) Calculate the design earthquake spectral response accelerations:
SDS
= 2/3 SMS
= 2/3(0.41) = 0.28g
SD1
= 2/3 SM1
= 2/3(0.28) = 0.19g
(4) For SDS
= 0.28g ASCE Table 11.6-1 shows that the SDC is B
For SD1
= 0.19g ASCE Table 11.6-2 shows that the SDC is C
Use the more severe of the two:
SDC = C
Considering the exception in ASCE 11.6 (see Section 11.2) check the building SDC:
Ts
= SD1
/SDS
= 0.19/0.28 = 0.68 seconds
Building T (see below) = 0.45 0.8(Ts
) = 0.8(0.68) = 0.54 seconds O.K.
The SDC can be determined based on Table 11.6-1 and the building assigned SDC B. For illustrative
purposes only this building example will continue using SDC C.
Table 11-1 shows that the building frame system with ordinary shear wall can be used for SDC C.
Fundamental Period
The approximate natural period of the structure can be calculated as follows:
T = Ct(hn)x
For building frame system Ct
= 0.02 and x = 0.75
T = 0.02(63)0.75
= 0.45 second (also T can be obtained from Fig. 11-8)

11-20
Seismic Response Coefficient:
For building frame system with ordinary shear walls, the modification factor R = 5 Table 11-1 (ASCE
Table 12.2-1). The occupancy importance factor I = 1 (ASCE 11.5).
Cs
should not be taken less than:
Cs
= 0.01
Cs
need not exceed:
Use Cs
= 0.055
Effective Seismic Weight
The effective seismic weight for this case includes the total dead load and the partition weight. The effective
seismic loads for different floors are calculated as follows:
For the first floor:
Slab = (121 ft)(61 ft)(0.142 ksf) =1048 kips
Interior columns (8 columns) = 8 (1.33 ft)(1.33 ft)(13.5 ft)(0.15 kcf) = 28.7 kips
Exterior columns (12 columns) = 12(1 ft)(1 ft)(13.5 ft)(0.15 kcf) = 24.3 kips
Walls = 2[(20.66 ft)(0.66 ft)+2(7.33 ft)(0.66 ft)](13.5 ft)(0.15 kcf) = 94.4 kips
Effective seismic weight (1st floor) = 1048 + 28.7 + 24.3 + 94.4 = 1195 kips
For second to fourth floor:
Slab = (121 ft)(61 ft)(0.142 ksf) =1048 kips
Interior columns (8 columns) = 8 (1.33 ft)(1.33 ft)(12 ft)(0.15 kcf) = 25.5 kips
Exterior columns (12 columns) = 12(1 ft)(1 ft)(12 ft)(0.15 kcf) = 21.6 kips
Walls = 2[(20.66 ft)(0.66 ft)+2(7.33 ft)(0.66 ft)](12 ft)(0.15 kcf) = 83.9 kips
Effective seismic weight = 1048 + 25.5 + 21.6 + 83.9 = 1179 kips
Cs =
SDS
R
I
⎛
⎝
⎜
⎞
⎠
⎟
=
0.28
5
1
⎛
⎝
⎜
⎞
⎠
⎟
= 0.055
Cs
=
0.19
5
1
⎛
⎝
⎜
⎞
⎠
⎟ 0.45
= 0.083
Cs =
SD1
R
I
⎛
⎝
⎜
⎞
⎠
⎟ T

11-21
For the roof:
Slab = (121 ft)(61 ft)(0.122 ksf) = 900 kips
Interior columns (8 columns) = 8 (1.33 ft)(1.33 ft)(6 ft)(0.15 kcf) = 12.8 kips
Exterior columns (12 columns) = 12(1 ft)(1 ft)(6 ft)(0.15 kcf) = 10.8 kips
Walls = 2[(20.66 ft)(0.66 ft) + 2(7.33 ft)(0.66 ft)](6 ft)(0.15 kcf)] = 42 kips
Effective seismic weight (roof) = 900 + 12.8 + 10.8 + 42 = 966 kips
The total effective seismic weight:
W = 1195 + 3(1179) + 966 = 5698 kips
Seismic Base Shear
V = Cs
W
V = 0.055 ϫ 5698 = 313.4 kips
Vertical Distribution of the Base Shear
The vertical distribution of the base shear V can be calculated using ASCE Eqs. 12.8-11 and 12.8-12
(Section 12.8.3). For T = 0.45 second the distribution exponent k = 1. The calculations for the lateral forces
and story shear are shown in Table 11-3:
Table 11-3 Forces and Story Shear Calculations
For building frame system the lateral force are carried by the shear walls. Gravity loads (dead load and live
load) are carried by the frames.
Distribution of the Seismic Forces to the Shear Walls in N-S Direction
For north south direction, consider that the lateral forces are carried by the two 248 in. by 8 in. shear walls. For
the symmetrical floor layouts the center of rigidity coincides with the center of mass at the geometric center of
the 120ft by 60ft floor area.
y
Level
Height
hx (ft)
Story
Weight
wx
(kips)
wxhx
k
Lateral
Force
Fx
(kips)
Story
Shear
Vx
(kips)
5 63 966 60,858 88 88
4 51 1,179 60,129 87 176
3 39 1,179 45,981 67 242
2 27 1,179 31,833 46 288
1 15 1,195 17,925 26 314
5,698 216,726 314∑

11-22
The shear force transmitted to each of the two shear walls must be increased due to code required displacement
of the center of mass (point of application of lateral force) 5 percent of the building plan dimension
perpendicular to the force direction. For the N-S direction the displacement considered = 0.05 ϫ 120’ = 6’.
To calculate the force in the shear wall the equation simplifies due toVi( )y
=
ki( )y
ki( )y∑
Vx
+
xi
ki( )y
Jr
Vx
ex
symmetry in geometry and element stiffness to for this case.
The shear forces and moments for the shearwall at each floor level are shown in Table 11-4:
Table 11-4 Shear Forces and Moments at each shearwall (N-S)
Load Combinations
For lateral force resisting (shear walls in this case) the following load combinations apply (Chapter 2 Table 2-6):
U = 1.4D Eq. (9-1)
U = 1.2D + 1.6L + 0.5Lr Eq. (9-2)
U = 1.2D + 1.6Lr
+ 0.5L Eq. (9-3)
U = 1.2D + 1.6Lr ± 0.8W
U = 1.2D ± 1.6W + 0.5L + 0.5Lr Eq. (9-4)
U = 1.2D ± E + 0.5L Eq. (9-5) (seismic)
U = 0.9D ± 1.6W Eq. (9-6)
U = 0.9D ± 1.0E Eq (9-7) (seismic)
The seismic load effect E in equations (9-5) and (9-7) includes the effect of the horizontal and vertical
earthquake induced forces (see Section 11.6.3). For SDC C the redundancy coefficient ρ = 1
E = QE + 0.2SDSD = QE + 0.2(0.28)D = QE + 0.056D For Eq (9-5)
E = QE - 0.2SDSD = QE - 0.2(0.28)D = QE - 0.056D For Eq (9-7)
Calculations of Gravity Loads on the Shearwall
Based on dead loads calculated in Chapter 6 Section 6.5.1
Vi( )y
=
1
2
Vx +
1
120
Vx ex
Level
Height
hx
(ft)
Lateral
Force
Fx
(kips)
Story
Shear
Vx
(kips)
Vxex
(kip-ft)
(Vi)y
(kips)
Moment
(kip-ft)
5 63 88 88 529.8 48.6 583
4 51 87 176 1053 96.6 1741
3 39 67 242 1454 133.2 3340
2 27 46 288 1731 158.7 5244
1 15 26 314 1887 173.0 7839

11-23
Total wall weight = 3.53(63) = 222 kips
Roof dead load = 0.122(480) = 59 kips
Four floors dead load = 0.142(480)(4) = 273 kips
Total dead load = 222 + 59 + 273 = 554 kips
Proportion total dead load between wall segments:
Two 8 ft segments: (2 ϫ 96 in.) = 192/440 = 0.44
One 20 ft-8 in. segment: (248 in.) = 248/440 = 0.56
For two 8 ft segments: Dead load = 0.44(554) = 244 kips
One 20 ft-8 in. segment = 0.56(554) = 310 kips
Roof live load = 20(480)/1000 = 10 kips
Four floors live load = (29.5 + 24.5 + 22.5 + 21) (480)/1000 = 47 kips (see Section 5.7.1)
Proportion of live load between wall segments:
Roof live load
For two 8 ft segments = 0.44(10) = 4.4 kips
One 20 ft-8 in. segment = 0.56(10) = 5.6 kips
4 floors live load
For 2-8 ft segments: = 0.44(47) = 21 kips
1-20 ft-8 in. segment = 0.56(47) = 26 kips
Recall wind load analysis (see Chapter 2, Section 2.2.1.1):
For wind load N-S direction
M = [(16.2 ϫ 63) + (31.6 ϫ 51) + (30.6 ϫ 39) + (29.2 ϫ 27) + (30.7 ϫ 15)]/2 = 2537 ft-kip
The axial force, bending moment, and shear acting on the base of the 20’-8” long shear wall resisting lateral loads
in the N-S direction are summarized in Table 11-5. Table 11-6 shows the factored axial force, bending moment and
shear. It is clear from the table that seismic forces will govern the design of the shearwalls in this example.
Story Drift and P-⌬ Effect
To check that the maximum allowable limits for story drift are not exceeded (ASCE Table 12-12.1) the
displacement at each floor level δxe need to be calculated and amplified (See Section 11.8). Then the stability
index θ should be calculated and checked against 0.1 where the P-⌬ effects are not required to be considered.

11-24
Table 11-5 Forces at the Base of Shear Wall N-S Direction
Table 11-6 Factored Axial Forces, Moment and Shear
Load Case Axial Force
(kips)
Bending Moment
(ft-kips)
Shear Force
(kips)
Dead load D 310 0
Roof live load Lr 5.6 0
Live load L 26 0
Wind W 0 2537 69.2
Earthquake E 0 7839 157
Load Combinations Pu
(kips)
Mu
(ft-kips)
Vu
(kips)
Eq. (9-1) U = 1.4D
434 0 0
Eq. (9-2) U = 1.2D + 1.6L + 0.5Lr
416.4 0 0
Eq. (9-3) U = 1.2D + 1.6Lr + 0.5L
393.96 0 0
U = 1.2D + 1.6Lr + 0.8W
380.96 2029.6 55.36
U = 1.2D + 1.6Lr - 0.8W
380.96 -2029.6 -55.36
Eq. (9-4) U = 1.2D + 1.6W + 0.5L+ 0.5Lr
387.8 4059.2 110.72
U = 1.2D - 1.6W + 0.5L+ 0.5Lr
387.8 -4059.2 -110.72
Eq. (9-5) U = 1.2D + E + 0.5L
385 7839 173
U = 1.2D - E + 0.5L
385 -7839 -173
Eq. (9-6) U = 0.9D + 1.6W
279 4059.2 110.72
U = 0.9D - 1.6W
279 -4059.2 -110.72
Eq (9-7) U = 0.9D + 1.0E
279 7839 173
U = 0.9D - 1.0E 279 -7839 -173

11-25
Design for shear:
The maximum factored shear force From Table 11-6:
Vu
= 173 kips
Determine φVc
From Table 6-5 for 8 in. wall φVc
= 7.3 x 20.67 = 150.9 kips
maximum φVn
= 36.4 x 20.67 = 752.4 kips
φVn
); however, shear reinforcement must be provided (Vu
φVc
).
Determine the required horizontal shear reinforcement
φVs
= Vu
- φVc
= 173 – 150.9 = 22.1 kips
φVs
= 22.1/20.67 = 1.1 kips/ft length of wall
Select horizontal bars from Table 6-4
For No.3 @ 18 in., φVs
= 2.6 kips/ft 1.1 kips/ft
Minimum horizontal reinforcement (of the gross area) for reinforcing bars not greater than No. 5, ρ = 0.002
(ACI 14.3.3) = 0.002(8)(18) = 0.29 in.2
area of No. 3
Use No. 4 @ 12 in spacing ρ = 0.2/(8 ϫ 12) = 0.0021
s = 12 in. smax
= 18 in.O.K.
Determine required vertical shear reinforcement (ACI Eq. 11-32):
ρ˜ = 0.0025 + 0.5(2.5 - hw/˜w )(ρh – 0.0025)
= 0.0025 + 0.5(2.5 – 3.05)(0.0021 – 0.0025)
= 0.0026
where hw/˜w = 63/20.62 = 3.05
Required Av
/s1
= ρ˜
h = 0.0026 ϫ 8 = 0.021 in.2
/in.
For No. 4 bars: s1
= 0.2/0.021 = 9.5 in. 18 in. O.K.

11-26
Design for flexure:
Check moment strength for required vertical shear reinforcement No. 4 @ 9 in
ρ˜
= 0.2/(9 ϫ 8) = 0.0028.
For 1-20 ft-8 in. wall segment at first floor level:
Consider Pu
= 279 kips
Mu
= 7839 ft-kips
˜w
= 248 in.
h = 8 in.
For No.4 @ 9 in.:
Ast
= 0.0028 ϫ 20.67(12)(8) = 5.5 in.2
Use No. 6 @ 6 in spacing ρ = 0.44/(6X12) = 0.0065
Ast = 0.0065 ϫ 20.67(12)(8) = 12.81 in.2
ω =
12.81
248 × 8
⎛
⎝⎜
⎞
⎠⎟
60
4
= 0.097
α =
279
248 × 8 × 4
= 0.035
c
w
=
0.035 + 0.097
2 0.097( )+ 0.72
= 0.144
ω =
5.5
248 × 8
⎛
⎝
⎜
⎞
⎠
⎟
60
4
= 0.042
α =
279
248 × 8 × 4
= 0.035
c
w
=
0.042 + 0.035
2 0.042( )+ 0.72
= 0.095
Mn
= 0.5 × 5.5 × 60 × 248 1+
279
5.5 × 60
⎛
⎝
⎜
⎞
⎠
⎟ 1− 0.095 /12 = 5699 ft − kips( )
φMn
= 0.9 5699( ) = 5129 ft − kips Mu
= 7839 ft − kips.

11-27
force transmitted to each of the four shear walls is for this case.
The shear forces and moments for each shearwall at each floor level are shown in Table 11-7:
Table 11-7 Shear Forces and Moments at each Shear wall (E-W)
For each two 8 ft segment:
Dead load = 244/2 = 122 kips
Roof live load for one 8 ft segments = 4.4/2 = 2.2 kips
Four floors live load for one 8 ft segments = 21/2 = 10.5 kips
From wind load analysis (see Chapter 2, Section 2.2.1.1):
For wind load E-W direction
V = (6.9 + 13.4 + 12.9 + 12.2 + 12.6) = 14.5 kips
M = [(6.9 ϫ 63) + (13.4 ϫ 51) + (12.9 ϫ 39) + (12.2 ϫ 27) + (12.6 ϫ 15)]/4 = 535 ft-kip
The axial force, bending moment, and shear acting on the base of the shear wall resisting lateral loads in the
E-W direction are summarized in Table 11-8. Table 11-9 shows the factored axial force, bending moment and
shear. It is clear from the table that seismic forces will govern the design of the shearwalls in this example.
Vi( )x
=
1
4
Vx
Distribution of the Seismic Forces to the Shear Walls in E-W Direction
For east west direction, the lateral forces are carried by the four 96” by 8” shear walls. The torsional stiffness
of the two N-S direction segments (248 in. each and 20 feet apart) is much larger than that of the four walls.
Assuming that the N-S walls resist all the torsion and neglecting the contribution of the E-W walls, the shear
Mn
= 0.5 ×12.81× 60 × 248 1+
279
12.81× 60
⎛
⎝
⎜
⎞
⎠
⎟ 1− 0.144( )/12 = 9267 ft − kips
φMn
= 0.9 9267( ) = 8340 ft − kips Mu
= 7839 ft − kips. O.K.
Level
Height
hx
(ft)
Lateral
Force
Fx
(kips)
Story
Shear
Vx
(kips)
(Vi)x
(kips) Moment
(kip-ft)
5 63 88 88 22 264
4 51 87 176 44 792
3 39 67 242 61 1518
2 27 46 288 72 2382
1 15 26 314 79 3560

11-28
Load Case Axial Force
(kips)
Bending Moment
(ft-kips)
Shear Force
(kips)
Dead load D 122 0
Roof live load Lr 2.2 0
Live load L 10.5 0
Wind W 0 535 14.5
Earthquake E 0 3560 79
Load Combinations Pu
(kips)
Mu
(ft-kips)
Vu
(kips)
Eq. (9-1) U = 1.4D
170.8 0 0
Eq. (9-2) U = 1.2D + 1.6L + 0.5Lr
164.3 0 0
Eq. (9-3) U = 1.2D + 1.6Lr + 0.5L
155.17 0 0
U = 1.2D + 1.6Lr + 0.8W
149.92 428 11.6
U = 1.2D + 1.6Lr - 0.8W
149.92 -428 -11.6
Eq. (9-4) U = 1.2D + 1.6W + 0.5L+ .5Lr
152.75 856 23.2
U = 1.2D - 1.6W + 0.5L+ 0.5Lr
152.75 -856 -23.2
Eq. (9-5) U = 1.2D + E + 0.5L
151.65 3560 79
U = 1.2D - E + 0.5L
151.65 -3560 -79
Eq. (9-6) U = 0.9D + 1.6W
109.8 856 23.2
U = 0.9D - 1.6W
109.8 -856 -23.2
Eq (9-7) U = 0.9D + 1.0E
109.8 3560 79
U = 0.9D - 1.0E 109.8 -3560 -79
Table 11-8 Forces at the Base of Shear Walls E-W Direction
Table 11-9 Factored Axial Forces, Moment and Shear (E-W direction)
Design for shear
The maximum factored shear force From Table 11-9:
Vu = 79 kips

11-29
Determine φVc
From Table 6-5 φVc
= 7.3 ϫ 8 = 58.4 kips
maximum φVn
= 36.4 ϫ 8 = 291.2 kips
maximum φVn
); however, shear reinforcement must be determined
Determine required horizontal shear reinforcement
φVs
= Vu
- φVc
= 79 – 58.4 = 20.6 kips
φVs
= 20.6/8 = 2.6 kips/ft length of wall
Use minimum shear reinforcement per ACI 14.3.3 similar to the 20 ft- 8 in. portion of the wall i.e. No.4
@12 in spacing (ρ = 0.0021). From Table 6-4 for No.4 @ 12 in.,
φVs
= 7.2 kips/ft 2.6 OK
Determine required vertical shear reinforcement
ρ˜
= 0.0025 + 0.5(2.5 - hw
/˜w
)(ρh
– 0.0025)
= 0.0025 + 0.5(2.5 – 7.88)(0.0021 – 0.0025)
= 0.0036
where hw
/˜w
= 63/8 = 7.88
Required Avn
/s1
= ρ˜
h = 0.0036 ϫ 8 = 0.029 in.2
/in.
For No. 6 bars: s1
= 0.44/0.029 = 15.2 in. 18 in. O.K.
Design for flexure:
Check moment strength for required vertical shear reinforcement No. 6 @ 15 in
ρ˜ = 0.44/(15X8) = 0.0037
For the 8 ft wall segment at first floor level:
Consider Pu = 109.8 kips
Mu
= 3560 ft-kips
˜w
= 96 in.
h = 8 in.

11-30
For No. 6 @ 15 in.:
Ast
= 0.0037 ϫ 96(8) = 2.82 in.2
In order to compensate for the big difference between φMn
and Mu
the thickness and reinforcement of this seg-
ment of the wall need to be increased. Increasing the thickness to 10 in. and use two layers off reinforcements.
Use two layers No. 6 @ 4 in. spacing
ρ = 2 ϫ 0.44/(10 ϫ 4) = 0.022
Ast
= 0.0022 ϫ 96(10) = 21.12in.2
ω =
21.12
96 ×10
⎛
⎝
⎜
⎞
⎠
⎟
60
4
= 0.33
α =
109.8
96 × 8 × 4
= 0.33
c
w
=
0.33+ 0.029
2 0.33( )+ 0.72
= 0.259
Mn
= 0.5 × 21.12 × 60 × 96 1+
109.8
21.12 × 60
⎛
⎝
⎜
⎞
⎠
⎟ 1− 0.259( )/12 = 4079 ft − kips
φMn
= 0.9 4079( ) = 3671ft − kips Mu
= 3560 ft − kips. O.K.
ω =
2.82
96 × 8
⎛
⎝
⎜
⎞
⎠
⎟
60
4
= 0.055
α =
109.8
96 × 8 × 4
= 0.036
c
w
=
0.055 + 0.036
2 0.055( )+ 0.72
= 0.109
Mn
= 0.5 × 2.82 × 60 × 96 1+
109.8
2.82 × 60
⎛
⎝
⎜
⎞
⎠
⎟ 1− 0.109( )/12 = 994 ft − kips
φMn
= 0.9 994( ) = 894 ft − kips Mu
= 3560 ft − kips.

11-31
11.11.2 Example 2 – Building # 1 Alternate (1) Standard Pan Joist
To illustrate the seismic design analysis for moment resisting frames Building 1 alternate (1) will be analyzed
for the north-south direction seismic forces.
Seismic design data
Assuming that the building is located in north Illinois with the maximum considered earthquake spectral
response accelerations for short period (0.2 second) and one second period determined from IBC maps
(2009 IBC Figures 1613.5(1) through 1613.5(14)) as follows:
Ss
= 0.20 g
S1
= 0.10 g
Soil site class definition is C
Occupancy category I
Determination of the Seismic Design Category SDC
(1) Based on the values of Ss
and S1
and soil site class C, determine the site coefficients Fa
and Fv
from ASCE Tables 11.4-1 and 11.4-2. Notice that linear interpolation is performed.
Fa
= 1.20
Fv
= 1.70
(2) Calculate the adjusted maximum considered earthquake spectral response acceleration:
SMS = Fa Ss = 1.20 ϫ 0.20 = 0.24g
SM1 = Fv S1 = 1.70 ϫ 0.10 = 0.17g
(3) Calculate the design earthquake spectral response accelerations:
SDS
= 2/3 SMS
= 2/3(0.24) = 0.16g
SD1 = 2/3 SM1 = 2/3(0.17) = 0.11g
(4) For SDS
= 0.16g ASCE Table 11.6-1 shows that the SDC is A
For SD1 =0.11g ASCE Table 11.6-2 shows that the SDC is B
Use the highest category from both tables
SDC = B
Considering the exception in ASCE 11.6 (see Section 11.2) check the building SDC:
Ts = SD1/SDS = 0.11/0.16 = 0.69 seconds
Building T (see below) = 0.43 0.8(Ts
) = 0.8(0.69) = 0.55 seconds O.K.

11-32
The SDC can be determined based on Table 11.6-1 and the building assigned SDC A. This will reduce the
seismic forces drastically. For illustrative purposes only this building example will continue using SDC B.
Table 11-1 shows that the building frame system with ordinary frame can be used for SDC B.
Fundamental Period
The approximate natural period of the structure can be calculated as follows:
T = Ct
hn
x
For moment resisting frame system Ct
= 0.016 and x = 0.9
T = 0.016(39)0.9
= 0.43 second (also T can be obtained from Fig. 11-8)
Seismic Response Coefficient:
For ordinary moment resisting frame system the modification factor R = 3 Table 11-1 (ASCE Table 9.5.2.2).
The occupancy importance factor IE
= 1 (2003 IBC Table 1604.85).
Cs should not be taken less than:
Cs = 0.01
Cs need not exceed:
Use Cs
= 0.053
Cs =
SDS
R
I
⎛
⎝
⎜
⎞
⎠
⎟
=
0.16
3
1
⎛
⎝
⎜
⎞
⎠
⎟
= 0.053
Cs
=
0.11
3
1
⎛
⎝
⎜
⎞
⎠
⎟ 0.43
= 0.087
Cs =
SD1
R
I
⎛
⎝
⎜
⎞
⎠
⎟ T

11-33
Effective Seismic Weight
The effective seismic weight for this case includes the total dead load and the partition weight. The effective
seismic loads for different floors are calculated as follows:
For the first and second floors:
Slab = (151.34 ft)(91.34 ft)(0.130 ksf) =1797 kips
Exterior columns (16 columns) = 16(1.34 ft)(1.34 ft)(13.0 ft)(0.15 kcf) = 56.0 kips
Effective seismic weight = 1888.1 kips
For the roof:
Slab = (151.34 ft)(91.34 ft)(0.105 ksf) = 1451.5 kips
Exterior columns (16 columns) = 16(1.34 ft)(1.34 ft)(6.5 ft)(0.15kcf) = 28.0 kips
Effective seismic weight (roof) = 900 + 12.8 + 10.8 + 42 = 1497.1 kips
The total effective seismic weight:
W = 1497.1 + 2(1888.1) = 5273.3 kips
Seismic Base Shear
V = Cs
W
V = 0.053 ϫ 5273.3 = 279.5 kips
Vertical Distribution of the Base Shear
The vertical distribution of the base shear V can be calculated using Eq. 12.8-11 (Section 12.8.3).
For T = 0.43 second the distribution exponent k = 1. The calculations for the lateral forces and story shear are
shown in Table 11-10:
Table 11-10 Forces and Story Shear Calculations
Level
Height
hx (ft)
Story
Weight
Wx
(kips)
Wxhx
k
Lateral
Force
Fx (kips)
Story
Shear
Vx (kips)
3 39 1,497 58,387 124 124
2 26 1,888 49,091 105 229
1 13 1,888 24,545 52 281
5,273 132,023 281∑

11-34
Story Drift and P-⌬ Effect
To check that the maximum allowable limits for story drift are not exceeded (ASCE Table 12-12.1) the
displacement at each floor level δxe
need to be calculated and amplified (See Section 11.8). Then the stability index
θ should be calculated and checked against 0.1 where the P-⌬ effects are not required to be considered.
Distribution of the Seismic Forces to the Frames in N-S Direction
For north south direction, the lateral forces are carried by the six frames shown in figure 1-3. For the symmet-
rical floor layouts the center of rigidity coincides with the center of mass at the geometric center of the 150’by 90’
floor area.
The seismic force transmitted to each frame must be increased due to assumed displacement of the center of
application of the applied lateral force (5 percent of the building plan dimension perpendicular to the force
direction). For the N-S direction the displacement considered = 0.05 ϫ 150ft = 7.5ft. To calculate the force in
each frame the equation can be used.
Assuming equal stiffness for the six frames the equation simplifies to:
simplifies to for this case.
where Fx
indicates the force for floor x
The seismic forces at each floor level are shown in Table 11-11:
Table 11-11 Shear Force at each Frame (N-S)
The axial, forces, shear forces, and bending moment for interior frame are calculated using the portal frame
method (see Chapter 2). The results of the calculations are shown in Figure 11-13
Jr
= x2
= 15( )∑
2
+ 45( )2
+ 75( )2
+ −15( )2
+ −45( )2
+ −75( )2
= 15750
Fi( )y
=
1
6
Fx
+
1
Jr
Fx
ex
Fi( )=
ki( )y
ki( )∑
Fx
+
xi
ki( )y
Jr
Fx
ex
ca
Level
Height
hx
(ft)
Lateral
Force
Fx
(kips)
Fxex
Frame
1
Frame
2
Frame
3
Frame
4
Frame
5
Frame
6
3 39 124 932.8 16.3 18.1 19.8 21.6 23.4 25.2
2 26 105 784.3 30.0 15.2 16.7 18.2 19.7 21.2
1 13 52 392.2 36.8 7.6 8.3 9.1 9.8 10.6

11-35
Figure 11-13 Calculations Results
Design of exterior span of the first interior beam N –S direction (first floor)
For live load and dead load calculations see Section 3.8.3
Dead load = 130(30)/1000 = 3.9 klf
Live load = 36.2(30)/1000= 1.1 klf
Moment and shear due to gravity loads may be calculated using the approximate coefficients from Figure 2-3
through Figure 2-7. The moment and shear due to wind loads were calculated in Chapters 2. The following
equations are used to determine the earthquake effects used for the ACI load combinations:
E = QE
+ 0.2 SDS
D = QE
+ 0.2(0.16)D = QE
+ 0.032D For ACI Eq (9-5)
E = QE
- 0.2S DS
D = QE
- 0.2(0.16)D = QE
- 0.032D For ACI Eq (9-7)
For the first exterior span
For ACI Eq. (9-5)
M = 104 + 0.032MDL
V = 6.9 + 0.032VDL
23.4 M = 25.4
V = 1.7
M = 25.4
V = 1.7
M = 25.4
V = 1.7
M = 72
V = 4.8
M = 72
V = 4.8
M = 72
V = 4.8
M = 104
V = 6.9
M = 104
V = 6.9
M = 104
V = 6.9
V = 3.9
M = 25.4
P = 1.7
V = 7.8
M = 50.8
P = 0.00
V = 7.8
M = 50.8
P = 0.00
V = 7.2
M = 46.7
P = 6.5
V = 3.9
M = 25.4
P = 1.7
V = 7.2
M = 46.7
P = 6.5
V = 14.4
M = 93.4
P = 0.00
V = 14.4
M = 93.4
P = 0.00
V = 8.8
M = 57.3
P = 13.4
V = 17.6
M = 114.6
P = 0.00
V = 17.6
M = 114.6
P = 0.00
V = 8.8
M = 57.3
P = 13.4
19.7
9.8

11-36
D L W E (ACI Eq 9-5) E (ACI Eq 9-7)
M (ft-kip) -199 -56.2 ± 99.56 ± 110.4 ± 97.6Exterior
support V (kips) 55.8 15.7 ± 6.64 ± 8.68 ± 5.1
Midspan M (ft-kip) 227.5 64.2
M (ft-kip) 318.6 -89.9 ± 99.56 ± 114.2 ± 93.8Interior
support V (kips) 64.3 18.1 ± 6.64 ± 9 ± 4.8
Exterior Support Midspan Interior SupportLoad Combinations
Mu
(ft-kips)
Vu
(kips)
Mu
(ft-kips)
Mu
(ft-kips)
Vu
(kips)
Eq. (9-1) U = 1.4D
-278.6 78.12 318.5 -446.0 90.0
Eq. (9-2) U = 1.2D + 1.6L + 0.5Lr
-328.7 92.08 375.7 -526.2 106.1
Eq. (9-3) U = 1.2D + 1.6Lr + 0.5L
-266.9 74.81 305.1 -427.3 86.2
U = 1.2D + 1.6Lr + 0.8W
-159.2 72.27 273.0 -302.7 82.5
U = 1.2D + 1.6Lr - 0.8W
-318.4 61.64 273.0 -462.0 71.8
Eq. (9-4) U = 1.2D + 1.6W + 0.5L+ 0.5Lr
-107.6 85.43 305.1 -268.0 96.8
U = 1.2D - 1.6W + 0.5L+ 0.5Lr
-426.2 64.18 305.1 -586.6 75.6
Eq. (9-5) U = 1.2D + E + 0.5L
-156.5 79.91 305.1 -313.1 95.2
U = 1.2D - E + 0.5L
-377.3 69.71 305.1 -541.5 77.2
Eq. (9-6) U = 0.9D + 1.6W
-19.8 60.84 204.8 -127.4 68.5
U = 0.9D - 1.6W
-338.4 39.59 204.8 -446.0 47.2
Eq (9-7) U = 0.9D + 1.0E
-81.5 55.32 204.8 -192.94 62.7
U = 0.9D - 1.0E -276.7 45.12 204.8 -380.54 53.1
For ACI Eq (9-7)
M = 104 - 0.032MDL
V = 6.9 - 0.032VDL
A summary for dead, live, wind and seismic loads moment and shear forces are presented in Table 11-12.
Table 11-12 Exterior Span Forces
Load Combinations
Table 11-13 shows the factored axial force, bending moment and shear.
Table 11-13 Factored Axial Forces, Moment and Shear

11-37
Design for flexure:
Check beam size for moment strength
Preliminary beam size = 19.5 in. ϫ 36 in.
For negative moment section:
where d = 19.5 – 2.5 = 17.0 in. = 1.42 ft
For positive moment section:
bw
= 20 (375.7)/172
= 26.0 in. 36 in.
Check minimum size permitted with:
bw
= 14.6(586.6)/172
= 29.6 in. 36 in. O.K.
Use 36 in. wide beam and provide slightly higher percentage of reinforcement at interior columns.
Top reinforcement at interior support:
From Table 3-5: Use 9-No. 9 bars (As
= 9.0 in.2
)
Check ρ = As
/bd = 9.0/(36 ϫ 17) = 0.0147 ρmin
= 0.0033 O.K.
Top reinforcement at exterior support:
As = 375.7/4(17) = 5.53 in.2
From Table 3-5: Use 7-No. 9 bars (As = 7.0 in.2
)
Check ρ = As/bd = 7.0/(36 ϫ 17) = 0.0114 ρmin = 0.0033 O.K
Bottom bars:
As = 375.7/4(17) = 5.53 in.2
= 6.32 in.2
)
As
=
Mu
4d
=
426.2
4 17( )
= 6.27 in.2
bw
=
20Mu
d2
=
20 586.6( )
172
= 40.6 in. 36 in.

Design for Shear :
Vu
at distance d from column face = 106.1 – wu
(1.42) = 97.0 kips
where
wu
= 1.2(3.9) + 1.6(1.1) = 6.4 klf
(φVc
+ φVs
)max
= 0.48 bw
d = 0.48(36)17 = 293.8 kips 97.0 kips O.K.
φVc
= 0.095 bw
d = 0.095(36)17 = 58.1 kips
φVc
/2 = 29.1 kips
Length over which stirrups are required = (106.1 – 29.1)/6.4 = 12.0 ft
φVs
(required) = 97.0 – 58.1 = 38.9 kips
Try No. 4 U-stirrups
From Fig. 3-4, use No. 4 @ 8 in. over the entire length where stirrups are required.
References
11.1 International Code Council, International Building Code, 2009
11.2 American Society of Civil Engineers, ASCE Standard Minimum Design Loads for Buildings and other
Structures, ASCE 7-05, 2005
11.3 S.K. Ghosh, and Qiang Shen, Seismic and Wind Design of Concrete Buildings (PCA LT276),
June 2008
11.4 ACI, Building Code Requirements for Structural Concrete (ACI 318-08) and Commentary (ACI 318R-02),
American Concrete Institute, 2002.
11.5 David A. Fanella, and Javeed A. Munshi, Design of Low-Rise Concrete Buildings for Earthquake Forces,
EB004, Portland Cement Association, 1998
11.6 David A. Fanella, Seismic Detailing of Concrete Buildings, SP382, Portland Cement Association, 2007
11-38

12-1
Chapter 12
Introduction to Sustainable Design
Sustainable development is thought to have three components: social, economic and environmental.
In the U.S., the Building Code Requirements for Structural Concrete (ACI 318-11, or the Code) addresses
the minimum design and construction requirements necessary to provide for public health and safety.
By protecting the public’s welfare, the Code meets the social implications of sustainable development.
Perhaps less obviously, the Code also allows for the environmental and economic components of sustainable
development. It permits use of environmentally responsible materials and provides durability requirements to
enhance lifecycle considerations in sustainable design. It also permits selection of structural systems to meet
the economic component of sustainable development and benefit the overall lifecycle evaluation of the entire
building. The introduction to ACI 318-11 clearly permits the owner or the Licensed Design Professional to
specify requirements higher than the minimums mandated in the Code. Such optional requirements can
include higher strengths, more restrictive deflection limits, and enhanced durability.
Ultimately, designing sustainably requires early interdisciplinary interaction of all stakeholders in the building
design and construction process, including the owner, architect, engineer, contractor, and material supplier.
Today’s best construction techniques and most advanced buildings are likely to become tomorrow’s standard
techniques and typical buildings. We continue to incorporate technological improvements in construction so
that we build better. We aspire to build the best that we can and hope that this saves money in the long run—
through longer lasting structures and more efficient operations.
Operating buildings requires a lot of energy; in the U.S., commercial and residential structures account for
about 40% of our total energy use, which is about 10% of the world’s energy use (Ref.12.1). Not only does the
Transportation
Industrial
Commercial
Residential
2008 2015 2020 2025 2030 2035
125
100
75
50
25
0
Figure 12-1 Primary United States’ energy use by end-use sector for 2008-2035, quadrillion
Btu. Commercial and residential buildings account for about 40% of the total energy demand.

12-2
built environment have a significant impact on energy consumption, it also affects water and material use
and waste generation (Ref.12.2). Therefore, what we choose to build and how we choose to build it have
increasingly important implications in a world where global climate change, growing populations, and rapid
depletion of limited natural resources (water and energy) are occurring simultaneously. Fortunately, we also
have more and better tools at our disposal to help make informed decisions about the buildings we build.
Simplified Design is primarily concerned with structural design in concrete; element sizing, optimized and
simpler reinforcement patterns, and good details. The concepts presented indicate ways to design buildings
more efficiently to save time, materials, or both. This introduction is intended to broaden the view from
structural implications to design considerations about materials and non-structural aspects. As buildings
become more complex, there are more considerations involved, and early planning becomes more even
important. By incorporating sustainable design attributes from the outset, it’s possible to maximize the
sustainable benefits.
Sustainable development and green construction include things like building layout and operations to mini-
mize operating energy, improve interior conditions, and advance the role of buildings in managing one of our
most precious resources: water. This chapter highlights various ways in which concrete can contribute
to sustainability, both as a material and as part of a building system. In this way, it is possible to speak
generically about using concrete to make a difference in a building’s performance and which attributes might
be used to advantage to demonstrate (and ultimately quantify) sustainability in various green rating systems.
This information is an introduction to a complex topic. The broad view does two things. First, it allows a
complete overview of a multifaceted topic—sustainable construction—that will continue to influence how we
build, perhaps increasingly so. Second, as the specific methods of achieving sustainable attributes evolve, the
broad view speaks to general goals rather than explicit choices. This helps prevent getting bogged down with
details of any particular rating system, such as earning points.
Key areas addressed by many of the green rating systems are as follows:
• Site selection and development
• System selection
• Water management
• Energy and atmosphere
• Materials and resources
• Indoor environmental quality
Following are some of the ways in which concrete systems can add to a project’s sustainability.
SITE SELECTION AND DEVELOPMENT
Concrete’s strength can be selected to meet the project needs. Higher strength allows us to build tall buildings.
This increases density, leads to smaller building footprints, and results in smaller amounts of material usage,
less habitat disturbance, and more open space. By keeping more vegetation, heat island effect is reduced.
This provides more comfort outside the building and reduces energy use to cool the buildings, which improves
air quality (indoors and outside) for everyone and saves building owners money.

SYSTEM SELECTION
Concrete is an economical, cost-effective solution, consuming minimal materials, energy, and other resources
for construction, maintenance, and rehabilitation over its lifetime, while providing essential infrastructure to
society. (Ref.12.3)
Concrete is strong in compression and has good stiffness (large elastic modulus). It can support heavy loads
and resist horizontal forces (wind, seismic) with only small amounts of deflection. This increases both the
safety and comfort of building occupants. If the concrete strength is increased, smaller sized members can be
used. Strength can be selected to reduce floor-ceiling heights to save on material quantities—exterior finishes
and glass/glazing, stair runs, etc. Due to its strength, durability, and solid construction, concrete is also a good
structural system for building green roofs.
WATER MANAGEMENT
Although Simplified Design is geared toward buildings, there are usually paved areas for walkways, parking,
and other uses that must be considered as part of the installation. As noted in the site selection information,
anything that decreases the amount of impervious surface on a site promotes better water infiltration into the
natural soil. Water that percolates into the ground finds its way into local waterways such as rivers, streams,
and lakes.
Two ways in which concrete can assist in managing water are by decreasing the size of the building footprint
and through the use of pervious pavements. Concrete products such as pervious concrete and permeable
interlocking concrete pavers (PICP) both promote recharging ground water. In addition to allowing natural
water to be replenished, pervious surfaces also improve water quality by helping to break down pollutants.
They may decrease the burden on a storm water system, reducing its size or eliminating it completely, with an
associated potential for saving money.
ENERGY AND ATMOSPHERE
This category generally reinforces the idea that buildings should not be energy drains. In fact, as we improve
the energy performance of buildings, we progress first toward minimal negative impact on the environment,
to net zero energy structures, to, ideally, buildings that help return energy to the grid. There are many exam-
ples of buildings being constructed today in each of these categories. The reason that concrete systems are
beneficial from an energy perspective is because they serve as useful structural elements and because they can
also help provide durable, energy efficient envelopes. Both the frame and the envelope have the potential to
provide thermal mass.
Energy codes are driving construction toward more efficient operations. With proper insulation strategies,
concrete systems create tight, energy efficient envelopes. The mass in these systems, which can act like a
“battery” to store heat or cool, reduces energy uses and improves occupant comfort. Depending on the
designer’s goals, it might be possible to expose some of the concrete frame—columns, walls, or floors—to
further enhance the thermal mass performance. And the building’s orientation and layout can be considered,
too. Exposing surfaces strategically to the sun can influence both the heating and cooling needs. Further, the
12-3
Chapter 12 • Introduction to Sustainable Design

12-4
orientation may affect the amount of daylight that illuminates the space versus the amount of artificial
lighting needed—and the way in which natural light contacts surfaces plays into thermal mass considerations.
MATERIALS AND RESOURCES
The key ingredient in concrete is cement. It is the glue that binds water and aggregates into a solid mass. While
cement manufacturing is energy intensive, improvements in the process over many decades have resulted in
green house gas (GHG) reductions per ton of material produced (Ref.12.5). And the bulk of concrete is
comprised of aggregates, which are very low embodied energy materials. Supplementary cementing materials
(SCMs), which are typically by-products of other industries, can replace a portion of cement to lessen
concrete’s environmental footprint.
In the U.S., fresh concrete is locally available most everywhere, with the average distance from a ready mix
plant to a project site being 14.4 miles (Ref.12.6). Most concrete ingredients are abundant, locally sourced, and
require minimal processing and transportation. Most materials used in the production of reinforced concrete
are also recycled or recyclable. As just noted, SCMs often come from other industries and using them in
concrete reduces landfill burdens. Crushed concrete can replace a portion of virgin aggregate in new concrete,
but it is more commonly used as base material for new pavement. Steel reinforcement (rebar)
is made predominantly from recycled rail steel. Recycling water from production and clean-out operations is
becoming more common in the ready mix industry, although it must meet certain chemical requirements
for use in concrete. Ultimately, reinforced concrete has a large potential for recycled content—and it can be
recycled at the end of a structure’s life, too.
Concrete systems for medium to taller buildings are especially desirable because the repetition allows for
formwork reuse. Climbing form technology offers speed and economy, which saves money for the building
and is resource efficient. It also implies that designers strive for repetition in building layout from one floor
to the next, which also leads to simplicity of design. And whether concrete is cast at the site or offsite,
panelizing the design can lessen material waste.
72°
Internal
External
Figure 12-2 Two beneficial effects of thermal mass are to reduce the peak interior temperature shifts
and to delay the time that the peak occurs.

An area that is not currently given much attention is the idea of how robust a structure is. Once a building
is in place, it is best to keep it functioning—because this maximizes the output we get from the materials
and resources invested in creating the structure. “The most sustainable building is the one that’s still
standing.” These ideas can be collectively referred to as “resilience,” which includes robustness, durability,
enhanced disaster resistance, and longevity (Ref.12.4). Concrete is strong to resist all kinds of forces, from
wind, flood, and earthquakes. And it is non-combustible so that it maintains structural integrity during a fire
to allow safe egress.
INDOOR ENVIRONMENTAL QUALITY
People spend about 90% of their time indoors. The aesthetic quality of an interior space plays a big role in
occupants’ well-being. Measures of the indoor quality include air, light/lighting, temperature, and noise.
Fresh air is important for everyone, and a small segment of society that is sensitive to chemicals has
additional special needs. Here concrete is beneficial because it is an inert material with no off-gassing.
Concrete’s structural capacity allows placement of openings as frequently as needed to provide good interior
day lighting throughout the space. And its mass is helpful for maintaining comfortable indoor temperatures
and providing privacy (sound attenuation) between different rooms.
There is not one “right way” to build sustainably. It is most effective to think of this as a process more than a
goal or an end point. This introduction has outlined some of the key areas and common topics addressed by
many of today’s green rating systems. We have described many ways that concrete is already being used to
advantage. New techniques will likely be discovered or invented as we move forward. Just as building codes
evolve as we learn from both innovations and failures, so, too, should sustainable design methods evolve to
incorporate the latest information that makes our buildings… simply better.
12-5
Chapter 12 • Introduction to Sustainable Design
Figure 12-3 Climbing forms result in more efficient construction.

References
12.1 EIA Annual Energy Outlook 2010, DOE/EIA-0383(2010), Environmental Information Administration U.S.
Department of Energy, Washington, D.C., May 2010. http://www.eia.gov/oiaf/archive/aeo10/index.html
12.2 Buildings and their Impact on the Environment: A Statistical Summary, Environmental Protection
Agency, Washington, D.C., April 2009. http://www.epa.gov/greenbuilding/pubs/gbstats.pdf
12.3 Design and Control of Concrete Mixtures, EB001.15, Portland Cement Association, Skokie, Illinois,
2011.
12.4 Protecting Lives and Property: Making Green Buildings “Functionally Resilient,” Portland Cement
Association, Skokie, Illinois, 2011.
http://www.edcmag.com/ext/resources/White_Papers/Saving-Lives-and-Property.pdf
12.5 2009 Report on Sustainable Manufacturing, Portland Cement Association, Skokie, Illinois, 2009.
www.cement.org/smreport09
12.6 2010 National Ready Mixed Concrete Association Fleet Benchmarking and Costs Survey, National
Ready Mixed Concrete Association, Silver Spring, Maryland, June 2010.
12-6

Simplified design of reinforced concrete buildings

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